A BRIEF HISTORICAL REVIEW ON FATIGUE

A BRIEF HISTORICAL REVIEW ON FATIGUE

A BRIEF HISTORICAL

OF CEMENTITIOUS LAYERS

REVIEW ON FATIGUE

I N ROAD STRUCTURES

CONTENTS

1.1

INTFODUCTION

1.2

HISTORICAL REVIEW

1.3

FAILURE CRITERION

1.4

OTHER RESEARCHERS

(1972 to 1976)

1.5

SUMMARY

1.6

DISCUSSION

1.7

REFERENCES

PAGE

1.1

1.2

1.9

1.10

1.14

1.14

1.15

Much work has been dune on the fatigue characteristics concrete and strongly cemented materials.

(Pretorius, of

1970;

Otte, 1972).

Most of the authors realised that cementitious layers in the road structures should be designed so that it not overstrained at the bottom; using the induced tensile

is

strain at the bottom of the layer as distress criterion, rather than tensile stress.

(Kaplan, 1963; otte, 1978).

This approach was mainly used to design cementitious base course layers for road structures.

When the "upside down" designs developed in the late 1970's, cementitious layers were used for the subbase layers.

During the period

prior

to 1979, relatively strongly cemented materials were used for the base and subbase layers.

Shrinkage or thermal cracking owing to excessive levels of drying shrinkage stresses, exceeding the tensile strength of the uncracked material, was intensi vely studied and reported on by various authors (Marais, 1973;

Groth, 1973; Biesenbach, 1973; Vail, 1973; Blight, 1973 and

Pretorius, 1973).

When these strongly cemented layers were used as subbase layers under relatively thick (100-200 mm) bitumen bases, reflection cracking caused great concern

(Freeme et al 1974, 1979, 1982).

These reflection cracks formed excellent paths for free wat~r ingress which eventually leads to rapid pavement deterioration (Otte, 1973).

During the late 1970' s and early 1980' s there was a strong move away from strongly cemented layers to unbound granular layers (Freeme, et al 1979).

It was however soon realized that the latter type of design resulted in relatively thick bitumen base layers being required to support the load.

Economical analyses (Freeme et aI, 1980) indicated however that bitumen base pavements with weakly* cemented subbase

The terms "strongly" and "weakly" are intended to reflect the initial 7-day strength (ucs) characteristic of the cemented layer whereas the terms "well" and "poorly" will be used to reflect workmanship or degradation.

layers can be cheaper than those with granular subbases.

This is accomplished by a reduction in the thickness of the bitumen base layer.

These new economic designs could save up to 25 per cent compared to the conventional designs of 1980 (Freeme et al, 1984).

It must be stressed that the strong move towards weakly cemented materials was mainly to overcome the drying shrinkage cracking, experienced with the relatively strongly cemented materials and not to reduce strength.

Proper definition of the difference between "weakly" and

"strongly" cemented materials is given in Chapter 2.

During the early 1980's a few experimental sections were constructed as well as normal high class traffic pavements in Natal using these economic designs.

Since that time, much of the national road structures included weakly cemented (lime, slag-cement or cement treated) subbase layers, usually 300 mm thick, built in two layers each 150 mm thick.

Recent investigations on the behaviour of cementitious subbase layers in South Africa, using the Heavy Vehicle

Simulator, (HVS) demonstrated that these subbase layers experience relatively early fatigue distress in the form of cracks.

The investigations indicates that these layers provide adequate support during the postcracked phase.

The

HVS tests indicated that for such pavements more than 80 per cent of the "life" of the cementitious subbase layers occurs in the postcracked phase.

In this chapter a short review is given on the fatigue characteristics of the earlier "strongly cemented" materials.

Pretorius (1970) studied and discussed "Design Considerations for Pavements Containing Soil Cement Bases" in the 1970's.

The most valuable part of his work, concerning this thesis, was the development of fatigue relationships and obtaining elastic and strength properties for strongly cemented materials.

"The fatigue characteristics of the soil-cement and the factors that influence its fatigue behav~our are therefore major factors to consider in the design of a pavement containing a soil-cement base".

Because of the importance of above statement, a fatigue investigation resulted in a fatigue-life versus strain relationship, illustrated in Figure 1.1.

The relationship between life (N) and initial maximum flexural microstrain (e.);

is:

J.

The relation can be used to indicate that flexural strains from 70 to 130 microns results in a fatigue-life spectrum of

10 to one million applications of load before any micro fatigue cracking is initiated at the bottom of soil-cement beam specimens.

See also Figure 1.1.

He further showed that the Poisson

I s ratio and volumetric micro strains for soil cement varies with stress level, between 0,1 and 0,5, and less than zero to 600, respectively.

Linear relationships between unconfined compression strength

(UeS) and tensile strength were also obtained.

There was, however a difference between the relationship obtained from direct tension and flexture tests.

The simple beam theory was used for his calculations, although he was aware of the fact that the modulus of elasticity in tension

is less than the modulus in compression (anisotropy).

The ratio of compressive to tensile flexural strength

is

larger than 5.

Pretorius mentioned also that the aggregate- mortar bond is the weak link in the strength of cement treated materials.

Other researchers (Jones, 1965; Taylor and Broms, 1964; HSU, 1963;

Kaplan, 1963; Shah and Winter, 1966), also reported in depth on this subject.

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FIGURE 1.1

FArlGUE LIFE OF FLEXURAL SPECIMENS

(PRErORIUS, 1970)

The modulus of elasticity and the unconfined compressive strength of the material tested by Pretorius for the fatigue relationship were 19300 MPa and 7MPa respectively, with a

A-1-0 non-plastic AASHO classification.

During the period 1972 to 1978, Otte worked in this field in

South Africa and proposed a fatigue relationship.

From literature this fatigue relationship was adopted and is now in use in the South African design method (Otte, 1978; Walker et aI, 1977; Freeme et al, 1982).

No fatigue work on cementitious materials was done by Otte.

The fatigue relation adopted is illustrated in Figure 1.2.

This relationship is: were Nf

=

Number of repetitions at strain £s to crack initiation.

It differs from the previous equation given by

Pretorius (eq 1.1), in that the strain at break, ~' is incorporated with the strain induced at crack initiation,

£.

Eq 1.1 incorporates only the initial maximum flexural s strain, £., which is comparable to £ in eq 1.2, without any

~ modification for shrinkage cracking (Freeme, et al 1982).

The modified strain is:

£

S

= d.

£

A comparison between the fatigue relationship suggested by

Otte, and the relationship obtained by Pretorius is given in

Figure 1.3.

It is important to note that eq 1.2 should be compared without the d-factor because Pretorius's result does not include factors for shrinkage in cemented material.

It is further important to note that a proper method to compare the two different relationships is to assume a

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(MODIFIED)

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STRAIN AT BREAK

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EQUIVALENT TRAFFIC.

E80 (Nf)

FIGURE 1.2

RECOMMENDED FATIGUE RELATION ADOPTED FOR

CEMENTITIOUS MATERIALS IN SOUTH AFRICA fFREEME,1982)

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FIGURE 1.3

COMPARISON BETWEEN THE FATIGUE RELATION

PROPOSED BY PRETORIUS (1970) 8 OTTE (978)

Factor, d, for modifying the tensile strain induced in cemented materials to allow for the presence of shrinkage cracking (Freeme, et al 1982) compressive strengths

(MFa)

Factor, d, for total thickness of cemented material (mm)

Weaklx Cemented:

Moderate cracking: crack widths less than 2mm

(eg natural materials with lime or 2-3\ cement

Stronglx cemented:

Extensive cracking: crack widths more than 2mm

(eg high quality natural gravels and crushed stone with 4-6\ cement)

0,75 1,5

1,5 - 3,0

1,1

1,15

1,2

1,3

The tensile strains at break, ~' for cementitious material with different unconfined compressive strengths are given in Table 1.2.

Tensile strain at break recommended for the standard cemented-material categories (Freeme, et al 1982)

Unconfined compressive strength (MFa)

Material type

Cemented at break

( ll£)

(\,)

6 - 12

3 6

1,5 3

0,75 - 1,5

Crushed stone (Cl)

Stone/gravel (C2)

Gravel (C3)

Gravel (C4)

120

125

145

strain at break for the Pretorius material, E:b

=

142pE:.

Equation 1.1 indicates that with initial maximum strain of

142 PE:,the fatigue life is zero.

Work done by Otte (1978), indicates however that for similar material to that of

Pretorius, an average strain at break of 145 pE: could be used.

Figure 1.3 indicates that the two relationships are consistant for induced strains greater than 100 PE, but for induced strains lower than 100PE the relationship suggested by Otte is more conservative.

Pretorius (1970) stated that "no simple failure criterion for the strength of soil-cement under general biaxial stresses exists at the moment".

Kaplan (1963) concluded that his test results on concrete suggested that the initiation of cracking may be more dependant on strain as distress determinant rather than on stress.

It was also suggested that engineers should develop a better understanding of l?train in general rather than stress, because strain values are more consistant than stress in cementitious materials (Otte 1972, 1978).

The traffic-induced strain concept was adopted in the South-

African design method in the 1970's (Walker et al, 1977), and is still in use, using eq 1.2.

Kekwick (1980) concluded in his work on strongly cemented soil slabs that:

"The general behaviour of the test slabs under load, together with the tensile responses of laboratory-scale specimens and a review of pertinent literature, has led to the adoption of a limiting tensile strain criterion as a fundamental material parameter regarding performance in a pavement system.

The limiting strain, defining the extent of elastic material response, --epresentsa maximum allowable deformation to which the material can be sUbjected in order to sustain many applications of load.

This is considered to be a fundamental characteristic of the concrete for highway application with the implication that induced tensile strains greater than the limiting value will lead to structural deterioration".

The author agrees with the above statements but according to the work given in this thesis it holds for less than approximately 20 per cent of the total life of a cementitious subbase layer.

According to HVS testing, which will be described later, more than 80 per cent of the "life" of a cementitious subbase layer occurs in the postcracked phase.

During this phase large blocks of cementitious material continue to crack untill the average block size reaches approximately the layer thickness.

When this happens the vertical compressive strength (resistance to shear) and erodibility of the cementitious material starts to govern the rut develoPment (permanent deformation on the surface).

The initial fatigue distress phase (precracked) does not contribute to any significant amount towards rut development.

Because "failure" of a road structure is defined mainly using terminal rut development criteria on the surface, the fatigue distress phase could only be seen as part of the structural rather than functional distress of a road structure, and occurs within 20 per cent of the total life of the structure.

The author agrees with Otte that fatigue distress can initiate functional failure through pumping.

It is however accepted that the initiation of permanent deformation on the road surface does not start with the first traffic-induced crack at the bottom of the cemented layers.

Other researchers: Raad (1976), Van Vuuren (1972), Marais

(1973) also worked in the field of cementitious materials.

The following section summarise their contributions briefly:

Raad (1976) applied the Griffith and modified Griffith theories to the failure of cement-treated materials.

He proposed a new approach to the fatigue life of cement-treated materials which incorporates the biaxial loading condition.

He proposed a relationship between the principal major and minor stresses

(CJ

1 and

"3) and the tensile strength of the

Note: + for compressive stresses

- for tensile stresses

He proposed a relationship between the stress level

FIT

and the number of load repetitions to failure (Nf) which is independent of the shape, width (time of loading or duration) and frequency of the applied stress pulses.

He applied the relationship to some published work on fatigue of cementtreated materials and it seems to fit the laboratory results obtained by several researchers much better than the simple relationship of stress or strain level versus the number of load repetitions.

The application of his laboratory study to pavement design still needs further evaluation, especially after the indication by Walker et al, (1977) that it does not really apply to the behaviour of actual pavements.

Using layered elastic theory and a strain criterion, Van

Vuuren (1972) calculated load equivalency factors for fracture in cement-treated materials and demonstrated that they where dependant on the structural layout of the pavement.

He calculated equivalency values ranging from 0,1 to 10 000 for cement-treated layers.

otte (1978) concluded that the existing concept of expressing an axle load spectrum as a certain number of equivalent 80 kN axles for a pavement with cement-treated layers is probably incorrect, not applicable and should preferably not be used.

It was suggested that each of the different wheel load intensities in the expected' traffic spectrum be accommodated when performing the design.

Since the number of expected load repetitions cannot be accommodated by calculating E80, some other way has to be devised to accommodate it.

It was suggested that Miner's cumulative damage law (Miner, 1945) should be assumed to be

valid and that a procedure similar to that developed by the peA (1966) for the design of concrete pavements be followed.

Marais (1973) reviewed testing and design criteria for cementtreated bases.

From his study it is concluded that unconfined compressive strength and durability tests should be done on these materials.

This is in accordance with a study in the

USA concerning suitable design criteria for cement-treated layers.

In 1967 the extent of use of various design criteria was as shown in Table 1. 3.

The table shows that the most popular criteria were 7-day unconfined compressive strength and durability.

A review by Williams (1973) of British experience concerning lean concrete roadbases, revealed that no durability criteria were used for their treated materials.

However, it is the author' s belief that because of the relatively high amount of cement used in lean concrete mixtures, excessive lack of durability is not a problem.

The durability problem arises mainly were "soil-cement" or weakly cemented materials are considered, especially fine grained materials.

The author believes that this should be investigated in depth in South

Africa.

Extent of use of different design criteria

(Marais, 1973)

Design Criteria Number of

States specifying

" of

Total

Durability criteria (both freeze-thaw and wet-dry tests) and unspecified compressive strength which increases with age and cement content (PCA criteria)

10

8 Durability criteria in addition to a definite 7-day unconfined compressive strength requirement

7-day unconfined compressive strength gene2 rally with occasional durability tests

7-day unconfined compressive strength in addition to freeze-thaw criteria only

3

Freeze-thaw durability criteria only in conjunction with increasing unconfined compressive strength

1

According to this review, most of the research on the behaviour of relatively strongly cemented layers in road structures, constitutes the fatigue characteristics under repetitive loading.

It is concluded that the main distress criterion

is

the maximum tensile strain at the bottom of such layers.

Biaxial stress conditions however were also investigated in the laboratory, but uncertainty exists in the applicability thereof in practice.

Investigations into the load equivalency of cementitious layers indicated a very wide range (0,1 to 10 000) for cement treated layers, and that equivalency should be incorporated in the design.

It was further shown interestingly enough that in the USAvarious design criteria during the 1960' s were used.

Of these the

7-day ues and durability were the most popular.

In South Africa, during the late 1970' s and early 1980' s cementitious layers were designed in accordance with its fatigue characteristics using the theoretical procedure, Otte

(1978)• Provision was also made for thermal stresses and shrinkage, by increasing the induced tensile strain at the bottom of the layer.

If the ratio between the modified induced strain

(E s) and the strain at break

(E b) are less than

0,3, the cementitious layer should at least withstand one million strain repetitions without fatigue distress.

(Otte,

1978) Extensive Heavy Vehicle Simulator (HVS) testing on weakly cemented subbase layers revealed relatively fatigue distress, afterwhich these layers early undergo advanced fatigue distress and break up into the postcracked phase.

In this phase, however the layer still provides adequate support in relatively dry conditions (Freeme et al, 1984).

The author considers that the behaviour of these materials in the postcracked phase must be incorporated into the design, including the rehabilitation design throughout the life of these types of pavements.

It will be shown later on that almost 90 per

cent of the "life" of the weakly cemented layers occurs in the postcracked phase.

Another very important aspect to consider with cementitious materials stabilized with lime, cement, lime/slagment etc., is the important difference between "modification" and "cementation" (Clauss, 1982).

It is however beyond the scope of this dissertation to discuss the rather complicated chemistry of soil cementation or modification.

The difference in "strength" and "durability" of cementitious materials is also important and this will be discussed in

Chapter 3.

PRETORIUS, P C (1970).

Design considerations for pavements containing soil-cement bases.

Ph.D

thesis, University of

California, Berkeley, California.

OTTE, E (1972).

treated materials.

The stressstrain properties of cement-

M.Sc. thesis, University of Pretoria (in

Afrikaans) •

KAPLAN, M F (1963).

Strains and stresses of concrete at initiation of cracking and near failure.

ACI Journal,

Proceedings V.60, No.7 pp. 8573-879.

OTTE, E (1978).

A structural design procedure for cementtreated layers in pavements.

DSc thesis, University of

Pretoria, South Africa.

MARAIS, G P (1973).

Transvaal Experience with Cement-treated crushed rock bases.

NIRR-PCI Symposium on Cement-treated

Crusher-run bases, Johannesburg, February, 1973.

GROTH, P

J

(1973).

The history and performance of Cementtreated crusher-run bases in Natal.

NIRR-PCI Symposium on

Cement-treated crusher-run bases, Johannesburg, February,

1973.

BIESENBACH, W J (1973).

Cement-treated bases in the Cape Pro-

~.

NIRR-PCI Symposium on Cement-treated Crusher-run bases, Johannesburg, February, 1973.

VAIL, J W (1973).

A review of the factors affecting the drying shrinkage cracking of cement-treated crusher-run bases.

NIRR-PCI Symposium on Cement-treated crusher-run bases,

Johannesburg, February, 1973.

crack patterns in roads with cement-treated bases.

NIRR-PCI

Symposium on cement-treated crusher-run bases, Johannesburg,

February, 1973.

PRETORIUS, P C (1973).

Shrinkage and traffic induced stresses in cement-treated crusher run bases.

NIRR-PCI Symposium on cement-treated crusher-run bases, Johannesburg, February,

1973.

FREEME, C R and MARAIS, C P (1974).

Traffic associated cracking of asphalt pavements.

Proc 2nd Conf.

on Asphalt

Pavements for Southern Africa, Durban, 1974.

FREEME, C R, and STRAUSS, J A (1979).

Towards the structural design of more economical pavements in South Africa.

Proc.

3rd Conf. on Asphalt Pavement for South Africa, Durban.

FREEME, C R, MAREE,

J

Hand VILJOEN, A W (1982).

Mechanistic

Design of Asphalt Pavements and Verification using the Heavy

Vehicle Simulator.

Proc. of the 5th Int. Conf. on Structural

Design of Asphalt Pavements, Delft, Holland, August 1982, vol

1, p 156-173.

Otte, E (1973).

The performance of two pavements containing cement-treated crusher-run bases.

NIRR-PCI Symposium on

Cement-treated crusher-run bases, Johannesburg, February,

1973.

FREEME, C R, OTTE, E and MITCHELL, M F (1980).

The economics of pavement type selection.

National Transport Commission Department of Transport, Pretoria, April, 1980.

FREEME, C R and WALKER, R N (1984).

Economic design of bituminous pavements.

Proc.

4th Conf.

on Asphalt Pavements for South Africa, Cape Town.

JONES, R (1965).

Cracking and Failure of Concrete Test

Specimen Under Uniaxial Quasi- Static Loading.

International

Conference on the Structure of Concrete and Its Behaviour

Under Load, Imperial College, London, September 28-30.

TAYLOR, M A and BROMS, B B (1964).

Shear bond strength between coarse aggregate and cement paste or mortar.

ACI

Journal, Proceedings V. 61, No.8

pp. 54-98.

HSU, T C (1963).

Mathematical analysis of Shrinkage Stresses in a Model of Hardened Concrete.

ACI Journal, Proceedings V.

60, No.3, pp. 371-389.

fracture of concrete.

American Concrete Institute, Publication SP-20, p. 5.

OTTE, E (1978).

A structural design procedure for cementtreated layers in pavements.

DSc thesis, University of

Pretoria, South Africa.

WALKER, R N, PATERSON, W DO, FREEME C R and MARAIS, C P

(1977). The South African mechanistic pavement design procedure.

Paper accepted by Fourth Int. Conf. on the Structural

Design of Asphalt Pavements.

to use for road construction.

Surrey, 1980.

RAAD, L (1976).

Design criteria for soil-cement bases.

Ph.D

thesis, University of California, Berkeley, California.

VAN VUUREN, D J (1972).

Discussion on paper by Brown and

~.

Proc.

Third Int. Conf.

on the Structural Design of

Asphalt Pavements, vol. II, p. 172.

MINER, M A (1945).

Cumulative damage in fatigue.

Applied Mechanics, vol. 12, September, p.A-159.

PCA (1966).

Thickness Design for Concrete pavements (ISOIO.

02P).

Portland Cement Association.

MARAIS, L R (1973).

Testing and design criteria for cementtreated bases.

NIRR-PCI Symposium on Cement-Treated Crusher-

Run Bases, Johannesburg, February.

WILLIAMS, R I T (1973).

Lean Concrete Roadbases - A review of

British Experience.

NIRR-PCI Symposium on Cement-treated

Crusher-Run Bases, Johannesburg, February.

CLAUSS, K A (1982).

Stabilization properties of dry processed, waste, carbide lime.

NITRR Report RS/2/82

(Unpublished), November, 1982.

ASPECTS OF THE CURRENT DESIGN METHOD OF

CEMENTITIOUS SUBBASE LAYERS IN SOUTH AFRICA

PRESENT DESIGN PHILOSOPHY FOR CEMENTITIOUS MATERIALS

IN SOUTH AFRICA

PRESENT DESIGN METHOD FOR CEMENTITIOUS LAYERS IN

SOUTH AFRICA

2.4.1 Shrinkagecracking

2.4.2 Fatigue and crack propagation

2.4.3 Cementitiouslayer in the equivalentgranular state

2.9

2.10

2.13

In order to understand the actual behaviour of road pavements, it is essential to define the necessary parameters controlling the behaviour of the pavement.

The concept of different states for the same structure during its service life was introduced.

Freeme (1984) defined the behavioural states of the pavement structures encountered in South Africa.

The author was involved in especially the developing of the understanding of the behavioural states of weakly cemented subbase layers under bitumen bases in Natal, South Africa.

It was found that these cementitious layers undergo marked changes in state during their behaviour as structural elements in the pavement.

It is therefore important to take this into account during the design stage.

As discussed in Chapter 1, a variety of economical pavements with asphalt bases was proposed during the late 1970's, based on mechanistic design principles, Freeme et al (1979).

During this period it was believed that better use could be made of the properties of asphalt bases by placing them over gravel subbases rather than strongly cemented subbases.

It was believed that less conservative fatigue criteria could lead to thinner asphalt bases.

According to Freeme (1984) significant advances were made in the design of bitumen base pavements.

According to economic comparisons made in 1980 (Freeme et aI,

1984) weakly cemented subbase layers appeared to be cheaper than granular subbase layers, based on the present worth of cost (PWOC) method.

Because of the economic implications especially for the high traffic class (E4)* roads, a number of experimental sections, based on the proposed economical designs were constructed during 1980 to 1982 in Natal.

The purpose of this dissertation, however, is not to discuss the economic implications but to investigate the behaviour of the

/36:2

/O..-b

/335

08b

weakly cemented subbase layers and their influence on the rest of the structure in terms of their structural and functional lives.

Suggestions of how to take account of the different behavioural aspects during the design and rehabilitation stage are made.

2.3

PRESENT DESIGN PHILOSOPHY FOR CEMENTITIOUS MATERIALS IN

SOUTHERN AFRICA

As discussed in Chapter 1, the method adopted in South Africa was to limit the horizontal tensile strain at the bottom of cementitious layers (strongly and weakly cemented) as proposed by Otte (1978).

More recently the concept of precracked and postcracked states for these layers was discussed by Freeme

(1984). From HVS testing and field behaviour of these layers, it was soon realized that most of the structural and functional life of pavements occur while the cementitious base- or subbase layers were in the postcracked state.

The behaviour of cementitious layers is best explained by initially discussing changes in its effective modulus which occur with time (traffic).

A typical example of changes in modulus a C2 quality material (See Table 2.1) will be discussed.

These changes are shown diagrammatically in Figure 2.1 (a)•

Initially in the precracked phase the effective modulus will be relatively high (3 to 4 GPa) and the layer will behave much as a slab of concrete,

i.e.

"very stiff" state.

However, shrinkage cracks and load associated cracking can reduce the effective modulus even though discrete large blocks of the material still retain the high modulus of the original cemented (stabilized) material.

The effective modulus can continue to drop to lower values in the order of 500 MPa, at which time the discrete blocks will be fairly small and could form a mosaic.

The behaviour at this stage is then very similar (equivalent) to that of high quality granular material but the structure changed into the "flexible" state.

The eventual modulus in this state will depend on the quality of

~ 4000

~

PHASE I

PRE CRACKED PHASE

I

I

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PHASE 2

POSTCRACKED PHASE

MODULUS DECREASES

BECAUSE Of CRACK I

PROPAGATION

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PHASE 3

INfLUENCE OF EXCESS

POREWATER PRESSURE

(EPWP)

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FIGURE 2.1

INDICATORS OF THE BEHAVIOUR OF CEMENTED

(C2 QUALITY) LAYERS ( AFTER FREEME, 1984)

the material originally stabilized, the cementing agent, the effectiveness of the mixing process, the absolute density achieved, the extent and degree of cracking and the sensitivity to excess porewater pressure (EPWP).

The ingress of water can significantly affect the modulus in this state through the development of excess porewater pressures (EPWP) and this stage has been designated Phase 3.

In some cases the layer may behave as a good-quality granular material with a modulus of 200 to 500 MFa, whereas in other cases the modulus will be as low as 50 to 200 MPa.

The net result of the cementitious layer reducing to these very low modulus values (50 MFa) will usually be inadequate support of any upper layers in the structure or inadequate cover over the lower layers or both.

These changes in modulus are reflected in the resilient behaviour of the cementitious material (see Figure 2.1 (b)•

Initially the relative deflection between the top and bottom of the slab could be small and will increase as cracking develops.

In Phase 3 the resilient deflection can increase markedly.

The plastic behaviour is also distinctive in that initially in Phases 1 and 2 there is little or no increase in deformation (see Figure 2.1(c).

However, in Phase 3, in the granular state, there can be a marked increase in deformation through densification, shear and/or erosion.

If the source of the water is removed and the material dries, its behaviour can again improve, (rate of change in deformation decrease).

In general surface resilient deflection is a good indicator of the state of cementitious layers; low def lections indicate high moduli and visa versa.

However, in Phase 3 the resilient deflection will not necessarily be a good indicator of large deformation, since this depends on the moisture sensitivity of the material (erodibility).

This however will be discussed in more detail in Chapter 3.

LONGER TIME DURING WHICH

MATERIAL BEHAVES AS A SLAB

(SHRINKAGE CRACKS CAN EXIST)

(/) 8

::>

..J

::> o

~

G

IJJ

>

~ 4 u

IJJ

IJJ

:t

2

FIGURE 2.2

SHEMATIC DIAGRAM OF RELATIVE BEHAVIOUR

OF CEMENTED LAYERS OF DIFFERENT fJUALITY fFREEME, 1984)

~

4

IJJ

>

~ u

IJJ u..

u..

IJJ

(/)

::>

..J

::>

0

0

~

3

2

I

0

C3

ORIGINAllY

STABiliZED

69 MATERIAL

TO C3 STANDARD

STATE

EGG (DRY)

EG9 (EXCESS POREWATER

PRESSURE)

FIGURE 2.3

SCHEMATIC REPRESENTATION OF THE EfJUIVALENT

BEHAVIOUR OF A CEMENTED MATERIAL fFREEME,1984)

The relative behaviour of cementitious materials of different qualities and strengths is indicated in Figure 2.2.

Generally high strength materials (such as a Cl)* will start off with a high modulus compared with that of a lower strength material

(e.g. a C4 material).

In the normal situation the higher strength materials will also tend to retain the high modulus state for a longer period.

Initially the cementitious layers will crack because of shrinkage and these cracks mayor may not be detrimental to the pavement.

However, when further cracking occurs, the rate of decrease of modulus is generally very rapid.

If measurements are made of the deflection during this period, when the decrease in modulus occurs the material will be found to be equivalent to that of lower strength cementitious materials (e.g. equivalent C2 quality, EC2).

The eventual equivalent state of the material is granular, upon breaking down into the smaller particle sizes (relative to layer thickness).

The quality of the equivalent granular material depends on the original granular material, the degree of densification achieved, the degree of interlock and particle strength. Generally, the higher the quality of material initially stabilized the higher will be the eventual equivalent modulus, if properly constructed.

The equivalent behaviour of a cemented material is shown diagranunatically in

Figure 2.3.

The effective moduli of cementitious materials are defined for four states : that is

(a) Before extensive cracking (precracked phase); although shrinkage cracks may and probably do exist the block size is large relative to layer thickness,t (block size >5t), and the material acts as a slab similar to a concrete slab.

In this state the layer should be designed against fatigue and the propagation of cracks.

*The codes for cemented materials are defined in Chapter 1, Table

1.2 (pl.B) and in Table 2.1 of this chapter.

(b) Loading or the environment, or both, have reduced the block size, but the behaviour is still predominantly controlled by the large blocks of material relative to layer thickness;

(St < block size < t). In this state the designer should again design against fatigue and further propagation of cracks. (See

Section 2.4.2)

(c) The block size has decreased to blocks small in relation to layer thickness, and the material is equivalent to that of a granular material (block size predominantly ~ t).

If the material

is,

in a dry state then the relevant safety factor (See Section 2.4.3) for the equivalent granular material should be used for design.

(d) The cementitious material has broken down into a granular state equivalent to (c), but the material

is

in a "wet" state.

The reduced moduli should be used for design, using also the safety factor approach.

The design phases which must be considered for this class of material can be divided into three phases.

These are

(a) Design against the effects of excessive shrinkage cracking.

(b) Design to control the effects of fatigue and propagation of cracks of the cementitious material while in the slab state.

(c) Design to control the possible shear deformation of the layer in an equivalent granular state.

TABLE 2.1 Moduli of cemented materials (Freeme, 1984)

Original code

UCS (MPa)precracked state

Parent material Precracked state

GPa (range)

Postcracked states (MFa)

Large Blocks Small Blocks

Dry s',ate Wet state

Equiv.

code

Equiv.

code

Cl

C2

C3

C4

6-12

3-6

1,5-3

0,75-1,5

Crushed stone G2

Crushed stone G3

14 (7-30)

12 (6-30)

Crushed stone G2

G3

Gravel G4

10 (4-14)

8 (14-14)

6 (3-12)

Gravel G4

G5

G6

G7

G8

5 (3-10)

4,513-9)

4 (2-8)

3,5(2-7)

3 (2-6)

G4

G5

G6

G7

G8

G9

G10

4 (2-7)

3,5(2-6)

3 (2-6)

2,5(1-5)

2 (1-4)

1,5(0,5-3)

1 (0,5-2)

Poisson's ratio 0,35

-

7-day UCS, Method A14, TH01,

(See ref. NITRR (1979».

3 000

2800

2 500

2 400

2 200

2 000

2 000

2 000

1 500

1 200

2 000

2 000

2 000

1 000

1 000

500

500

600

600

EG1

EGI

500

450

450

EGI

EG2

EG2

400

350

300

250

200

EG3

EG4

EG4

EG5

EG5

350

300

250

200

170

150

125

EG3

EG4

EG4

EG5

EG5

EG6

EG6

500

400

EGI

EG2

300

250

200

EG2

EG3

EG3

180

160

140

120

90

EG4

EG5

EG6

EG7

EG8

180

160

140

120

90

70

45

EG4

EG5

EGG

EG7

EGO

EG9

EG10

(X)

.

N

I

Factor, d, for modifying the tensile strain induced in cemented materials to allow for the presence of shrinkage cracking (Otte, 1978) compressive strengths

(MFa)

Factor, d for Total thickness of cemented material (mm)

Weakly cemented:

Moderate cracking; crack widths less than 2 mm

(e.g. natural materials with lime or 2-3 , cement)

Strongly cemented:

Extensive cracking; crack widths more than 2 mm

(e.g. high-quality natural gravels and crushed stone with 4-6 , cement)

0,75 - 1,5

1,5 3,0

1,1

1,15

1,2

1,3

Shrinkage cracking : Cemented (strongly and weakly cemented) materials generally crack through shrinkage as a result of drying and thermal stresses.

This is taken into account in design by increasing the traffic-induced strain,

E, by a factor, d, which can be obtained from Table 2.2.

The increased value is termed (after Otte, 1978) the modified strain,

E , viz.: s e: s

= d •

e: •••••••••••••••••••••.•••••••••••••••••

(2.1)

below the cemented layer must also be given consideration.

For the layer above the activity of the crack

is

of importance since high movements (activity) may be present at the crack.

At present, research is being done to determine not only the movements at cracks but also how to take account of this effect for rehabilitation design, Freeme

(1984) •

The maximum tensile strain at the bottom of the layer

is

adopted as the criterion for controlling cracking under traffic.

The material is very sensitive to the magnitude of tensile strain and has a limited range of fatigue life,

(precracked phase).

The equivalent traffic required to cause initiation of cracking

is

determined from Figure 2.4.

This requires computation of the strain ratio (£s/£b)' where £b

is the tensile strain at break given in Table 2.3.

C1

C2

C3

C4

Code

Tensile strain at break recommended for the standard cemented-material categories

Cemented material:

UCS (MPa)

Material type cemented

Tensile strain at break £b(lJ£)

6-12

3-6

1,5-3

0,75-1,5

Crushed stone

Stone/gravel

Gravel

Gravel

145

120

125

145

Fatigue properties : The fatigue life of relatively strongly· materials under repeated flexure can be expressed by the following

This relationship is also adopted for weakly cemented i.e. C3 and C4 materials in the precracked state.

J:l

\II

1,0 z e:t

Q;

~

(J)

"-

\II

II'>

0,8

0

~ e:t

Q;

0.6

0.4

0,2

ESIGNTRAFFIC CLASS Eol EI EZ

..

I

E3

I

E4

III

.• I

11111 ••

1=

ES

=

TRAFFIC INDUCED STRAIN (MODIFIED FOR

~l

~Eb

TH.E EFFECT OF SHRINKAGE CRACKS)

=

STRAIN AT BREAK

-

I

•...

o

103

10

4

10~ 10

6

ECUIVALENT TRAFFIC, E801 Nf)

10

7

(USE RELATIVE DAMAGE COEFFICIENT OF d : 6

10

8

FIGURE 2.4

RECOMMENDED STRAIN RArlO FOR CEMEN'TED

LAYERS CRACK INITIATION-

z e:t

Q;

~

(J)

J:l

\II

0,8

"-

II'>

\II

0.6

Q

~

~ 0,4

0,2 o

10

10 10 10

EOUIVALENT TRAFFIC, E80

(USE RELATIVE DAMAGE COEFFICIENT OF d : 6 )

FIGURE 2.5

ALLOWABLE STRAIN RATIOS FOR CEMENTED

LAYERS.

- VISIBLE CRACKS AT THE SURFACE-

10

9,1 (1 where Nf

= number of repetitions at strain €s to crack initiation.

A shift factor is applied to account for time between initiation of cracking and cracks appearing at the surface.

Table 2.4 shows the applicable shift factors for the different road categories and for different layer thicknesses while from Figure 2.5, the equivalent traffic required before cracks appear can be determined directly from the strain ratio.

In order to convert different axle loads to the standard E80, it is recommended to use a relative damage coefficient, d

=

6, in the precrackedphase*.

It must be emphasized that the precracked phase is the phase where the slabstate of the cementitious layer controls the behaviour of the pavement.

Layer thickness

(mm)

Shift factors** for cemented layers

(Freeme, 1984)

To convert the number of load repetitions other than 40 kN

(80kN axle) the relative damage formula, F

=

(4~)d is used, with d

=

6.

The calculated fatigue life is multiplied by the appropriate shift factor

A cemented layer can crack under the action of traffic into smaller particle sizes relative to the layer thickness.

Often there is still very good interlock between the particles and effectively the material behaves as an equivalent granular material, even though it will not meet the specification for the granular material.

According to Freeme (1984), the criteria applicable to granular materials should apply for granulated cementitious materials.

It is worthwhile to discuss the distress criterion for granular materials in the following section, as it is proposed to apply to cementitious materials in the equivalent granular state.

An important concept here

is

that

if

a material

is

in an equivalent material state, then the criteria for that state apply, i.e.

if cementitious material is in the granular state, say EG4 state, then the approach is to use safety factors to control deformation of the material (e.g. safety factors and Mohr-Coulomb strength parameters of cohesion, c, and angle of internal friction, ~, for a G4 material apply,

Maree (1982)).

Granular materials exhibit distress in the form of cumulative permanent deformation or inadequate stability.

Both distress modes are related to its shear strength which is expressed in terms of the Mohr-Coulomb strength parameters of cohesion, c, and angle of internal friction, ~.

The use of the safety factor, F*, as defined

in

the equation (Maree 1982) below, safeguards the layer against shear failure or gradual shear deformation by limiting the shear stresses to a safe level.

______ 11..-_____ _

A _

IK(tan2 (45 + ~/2)-~1 + ;Kc tan (45+ ~/:)

(0'1 0'3)

where 0'1and 0'3

= calculated major and minor principal stresses acting at a point in the layer.

(Compressive stresses are positive while tensile stresses are negative.) c

= cohesion

~ = angle of internal friction

K

= constant

=

0,65 for saturated conditions

=

0,80 for moderate conditions

=

0,95 for normal dry conditions

The values of c and ~ and the c-term and ~-term in eq. 2.3

are given in Table 2.5.

Values are given for the material in the dry state (~ 45 per cent of saturation) and the wet state (~

90

Per cent of saturation).

According to eq. 2.3 the safety factor depends on 0'1 and

0'3.

These two stresses dePends more on contact pressure and contact area, than on wheel load, thus is F indirectly affected.

The safety factor depends therefore more on the contact pressure and it

is

therefore recommended that a

40 kN dual wheel at

520 kPa contact pressure be used.

The safety factors should be calculated at the mid-depth of the granular layer, under one of the wheels and at the centre of the dual wheels.

The safety factor, F, must not be mistaken for the "F" of the relative damage formula discussed on page 2.12.

Moisture

State

High density crushed stone, Gl dry wet

Moderate density crushed stone, G2 dry wet

Crushed stone and soil binder, G3 dry wet

Base quality gravel, G4 dry wet

65

45

45

35

Subbase quality gravel,

G5

Moderate wet

Low quality

Moderate subbase gravel, G6 wet

40

30

30

25

55

40

50

35

Friction, 0

(Degrees)

55

55

52

52

50

50

48

48

43

43

40

40

8,61

5,44

392

171

7,06

4,46

303

139

6,22

3,93

261

115

5,50

3,47

223

109

3,43

3,17

147

83

2,88

1,76

103

64 recommended safety factors for the design traffic classes

(also shown in Figure 2.6).

For cementitious materials in the equivalent granular state of G7 to GIO, the material should be treated as an equivalent subgrade. In cases where a lens of the material

is

sandwiched between two stiff layers, the safety factor approach is desirable for analysing the lens.

At present, values for c and ~ for materials of this standard have not been determined but estimates are given in Table 2.7.

DESIGN"

TRAFFIC

CLASS

2,0

4

U) tL.

0::

0

~

0 u...

>-

~

1,0

A

B o

104

C

".

"

EO

-

I I

EI

E2~ E3

-E4-

ROAD

CATEGORY: -

A-

B:

.--

...,

".

••• l..oo

C-

-

.-

~

,.-

.-""

•••••

••• i"'"

~

--

...,

-

-

-

10~ 10

6

107

EOUIVALENT TRAFFIC, E80

RECOMMENDED VALUES OF RELATIVE DAMAGE COEFFICIENT

TO COMPUTE EOUIVALENT TRAFFIC d- 4 FOR A CATEGORY ROAD d= 3 FOR B CATEGORY ROAD d= 2 FOR C CATEGORY ROA D

FIGURE 2.6

RECOMMENDEDSAFETY FACTORS FOR GRANULAR

M~TERIALS fGf TO "G6) fFREEME, 1984)

Road

Category

A

B

C

E4

E3

E3

E2

E1

E2

E1

EO

The safety factor recommended for granular material (Maree, 1982)

Design Traffic

Class

Minimum Allowable

Safety Factor

1,60

1,40

1,30

1,05

0,85

0,95

0,75

0,50

Estimates of values of c and ~ for soils

(G7 to G10)

Material

Code

Moisture

State Friction,

~ (Degrees)

G7

G8

G9

G10 dry wet dry wet dry wet dry wet

25

20

25

20

25

20

25

20

35

35

30

30

28

28

25

25

2,51

1,40

1,85

0,95

1,63

0,80

1,34

0,60

79

43

75

41

91

50

82

45

Although the concept of equivalent granular states for cementitiuos layers was derived from the findings of HVS research on cementitious layers in road structures, the actual quantification of all the necessary components to result in accurate predicted behaviour for these layers had not been done prior to this dissertion.

The proposed method to use assumed c and

~ values for the equivalent granular states, however, applied with different levels of success by different people and organisations.

A suggested way to improve the validity of the concept is to analyse the HVS results in such a format that pavement behaviour and behavioural states can be directly br indirectly predicted from field results.

However the South

African design method is unique in utilizing the concept of different states of the pavement during its service life.

The main reason for this advantage in design is the availability of the HVS.

An attempt will be made by the author to further evaluate the above-mentioned concept using recent HVS and laboratory results on weakly cemented materials.

(Lime treated) These tests were conducted by the author during the period January 1983 to 1985.

Most of the HVS tests had been done on asphalt base and cementitious subbase pavement structures in Natal on National routes, N2 and N3.

These tests were carried out on four different structures with basically the same initial design.

In order to understand the behaviour of the cementitious subbase layers it is therefore necessary to investigate these layers not as separate structural units only, but to study the influences of these layers on the rest of the structural layers in the total system (structure), inclUding the asphalt layers. The reason for this approach is that the different pavement layers behave as a system and the behaviour of each layer is dependant on the behaviour of the layer above or below.

The behaviour of the cementitious subbase layers in different states will be frequently highlighted.

It is believed that the upper subbase layer, in this design,

is

the most critical structural layer and mainly governs the performance of the structure.

This will be evaluated and discussed further in Chapter 3.

The effect of strength build-up because of the cementing action with time as well as factors such as the effect of delayed compaction to reduce cracking were not evaluated during this study.

It is however accepted that these factors are very important and should be investigated further in the future.

This includes the durability of these materials.

FREEME, C R (1984).

Symposium on Recent findings of Heavy

Vehicle Simulator testing.

ATC 1984, NITRR, South Africa.

FREEME, C R and STRAUSS,

J

A.

(1979).

Towards the Structural design of more economical pavements in South Africa.

Proc. of the third Conf.

on Asphalt Pavements for southern Africa,

Durban, Vol.l.

FREEME, C R and WALKER, R N.

(1984).

Economic design of bituminous pavements.

Proc.

of the Fourth Conf.

on Asphalt

Pavements for southern Africa - CAPSA84, Cape Town, Vol 1.

Structural design of interurban and rural road pavements.

Pretoria, TRH4, CSIR.

OTTE, E (1978).

A structural design procedure for cementtreated layers in pavements.

DSc thesis, University of

Pretoria, South Africa.

NATIONAL INSTITUTE FOR TRANSPORT AND ROAD RESEARCH (1979).

Standard Methods of Testing Road Construction Materials.

Technical Methods for Highways, No.1, CSIR, Pretoria, RSA.

1979, ix

+ 183 pp.

MAREE,

J

H (1982).

Aspekte van die ontwerp en gedrag van

Padplaveisels met korrelmateriaalkroonlae.

D.Sc. Dissertation

(in Afrikaans).

University of Pretoria, Pretoria.

EVALUATION AND ANALYSES OF A NUMBER

OF HEAVY VEHICLE SIMULATOR TESTS AND RESULTS

SUMMARY OF HVS TESTS AND RESULTS AT MARIANNHILL,

N3/1

3.2.1

3.2.2

3.2.3

3.2.4

3.2.5

3.2.6

3.2.7

3.2.8

3.2.9

3.2.10

3.2.11

3.2.12

3.2.13

3.2.14

3.2.15

Road structure and subbase material

Selection of HVS test sections

Permanent deformation

Transverse surface profiles

Road surface deflection (RSD) and radii of curvature (RC)

Resilient depth deflections

Mechanistic analyses

Layer densities and moisture contents

Permeability test results (MARVIL)

Dynamic Cone,Penetrometer (DCP) test results

Block sizes of cracked weakly cemented subbase material

Summary of some indicators of behaviour

Pavement state and radius of curvature

Development of a resilient behavioural model

Summary and conclusions of HVS tests at

Mariannhill, N3/1

3.3

3.4

3.7

3.20

3.20

3.27

3.29

3.39

3.42

3.46

3.48

3.49

3.51

3.56

SUMMARY OF HVS TESTS AND RESULTS AT FIGTREE,

N2/24

3.3.1

3.3.2

3.3.3

3.3.4

3.3.5

3.3.6

Road structure

HVS test programme

Permanent deformation

Road surface deflection (RSD) and radii of curvature (Re)

Resilient depth deflections

Summary and conclusions of HVS tests at

Figtree, N2/24

3.70

3.70

3.70

3.75

3.77

SUMMARY OF HVS TESTS AND RESULTS AT UMGABABA,

N2/24

3.4.1

3.4.2

3.4.3

3.4.4

3.4.5

3.4.6

Road structure

HVS test programme

Permanent deformation

Road surface deflection (RSD) and radii of curvature (RC)

Resilient depth deflections

Visual observations and discussions

SUMMARY OF HVS TESTS AND RESULTS AT VAN REENEN' S

PASS, N3/6

3.80

3.82

3.83

3.86

3.86

3.5.1

3.5.2

Section 1

Section 2

Observation of fatigue distress of the asphalt layers

3.91

3.93

3.5.4

3.5.5

Road surface deflection (RSD) and radii of curvature (RC)

Resilient depth deflections

Mechanistic analyses

3.5.5.1

3.5.5.2

Effective elastic moduli

Prediction of future expected fatigue distress

Validation of predictions and discussions

3.6

3.7

SUMMARY AND DISCUSSIONS

REFERENCES

3.102

3.105

3.111

3.116

3.124

3.127

3.130

Prior to this effort of full scale research, not much work has been done to quantify the behaviour of weakly cemented layers in road structures.

The current South African mechanistic design method heavily rely upon the fatigue relationship proposed by Otte (1978) mainly in the precracked phase.

With the use of the HVS the limitations of the design and theoretical evaluation method was soon realized.

The concepts of precracked and postcracked and also equivalent granular state of cementitious layers (Freeme et aI, 1984) are direct results of HVS research.

The purpose of this chapter is to describe the behaviour of weakly cemented subbase layers including its influence on the other layers (surfacing, base and subgrade) in the road structure.

Four (4) HVS tests, done by the author during the period January 1983 to December 1984, will form the basis of evaluation and discussion.

The research include mainly work done on bituminous base structures (Freeme, 1979,

1984) found in the Province of Natal, South Africa.

The basic pavement design specifies a 40 romasphalt surfacing, bitumen base (80-135 romthick) on two weakly stabilised subbase layers, each 150 romthick.

Beneath the subbase layers the selected subgrade (150-300 mm thick) and the subgrade occur.

The four sites tested with the HVS included in this chapter are at Mariannhill, National Route 3, Section 1

(N3/1)i Figtree, National Route 2, Section 24 (N2/24)i

Umgababa, National Route 2, Section 24 (N2/24) and Van

Reenen's Pass, National Route 3, Section 6 (N3/6).

Reference will also be made to other related HVS tests and findings.

In this thesis only a comprehensive summary is given of the tests and results of above-mentioned tests.

The detailed descriptions and evaluations of each test are given elsewhere (De

Beer, 1984{a) ,(b)i 1985(a),{b».

Photographic records of these HVS tests are given in Appendix A

Three separate HVS tests were done at this site.

These tests include dry tests and water introduction into the sublayers and also on one section the asphalt was artificially heated for specified periods during the testing.

The state and condition of each layer in the structure were monitored before, during and after testing.

Resilient deflections and permanent deformations of each layer were measured with the multi-depth deflectometer (MOD).

Surface profiles across the test sections were measured using a straight edge and electronic profilometer.

Surface deflections were measured with the road surface deflectometer (RSD), which is a modified

Benkelman beam (De Beer, 1984(a».

The road structure at Mariannhill is shown

in

Figure 3.1.

The design specified 40

IlIIlI

AC; 125

IlIIlI

BC; 150

IlIIlI

C3;

150

IlIIlI

C4; 250

IlIIlI

G4; G7 in situ material.

These material codes are according to TRH14 (NITRR, 1985).

The variation in layer thicknesses are also shown in the figure.

Both the upper and lower weakly stabilized subbase layers consists of wheathered granite and were classified using the

Highway Research Board's (HRB) classification system as

Al-b(O) before stabilization.

Lime (Ca(OH)2) was used for the stabilization.

In the upper subbase an average of 2,4 per cent (by mass) and in the lower subbase an average of

2,1 per cent lime were added.

The parent plasticity (before stabilization) varied between semiplastic and 5 per cent.

Average relative compactions of 97,6 and 96,8 mod. AASHTO were obtained for the upper and lower subbase, respectively.

The unconfined compressive strengths (UCS) after 24 hours curing at 70 to 75°C were 2,4 MFa and 2,5 MPa for the upper and lower layer, respectively.

Minimum UCS's of 1,5 MFa and

0,75 MFa were specified for the upper and lower layer.

According to these results both layers could be classified

• LAYER THICKNESS

ACCORDING TO TRHI4

AND MATERIAL

VARIATION IN

THICKNESS lmm)

35 - 40

110-125

CODE

CONTINUOUSLY GRADED

ASPHALT PREMIX SURFACING

CONTINUOUSLY

BITUMEN HOT-

GRADED

MIX

YELLOW DECOMPOSED

GRANITE

+

3-/.

LIME

IUCS~2,4 MPo)

DARK YELLOW ORANGE

DECOMPOSED GRANITE

LIME lUCS~2,5 MPo)

+

3-/.

REDDISH BROWN DECOMPOSED

SANDSTONE ICBR

<:

80; P.I ~6)

FIGURE 3.1

PAVEMENT STRUCTURE AT MARIANNHILL HVS

. SITE

using the material codes in TRH14 (NITRR, 1985) as C3material (UCS: 1,5 MPa to 3,0 MFa).

The base layer consisted of 50/50 Pen. hot-mix asphalt (continuously graded) and the selected and subgrade layers of decomposed sandstone.

Selection of HVS test sections

Prior to HVS testing an extensive road surface deflection survey was conducted and was reported elsewhere, De Beer

(1984(a».

The survey indicated that the deflection ranges between zero and 0,21 mm, with an average site deflection was 0,12 mIllwith a standard deviation of

*

The locations of the three selected test sections as well as the deflections are illustrated in Figure 3.2(a) and (b).

An important observation gained from the deflection survey was that although the deflections on the selected subgrade were relatively low (0,8 mm) during construction, the final deflections on the surface were reduced to an average of

0,12 mm.

The construction of 2 x 150 mm thick weakly cemented subbases, 1 x 125 mIllthick bitumen base and 40 mIll thick asphalt surfacing, were seen to reduce the deflection on the selected subgrade by approximately seven times, one year after construction.

This is a relatively good illustration of the design principle covering the subgrade with successive layers to reduce stresses (deflections) on the subgrade layers.

It is believed that the cementitious properties of the weakly cemented subbase layers are the main contributor to this reduction in deflection.

An important question concerning the permanency of these cementitious properties with time and or traffic loading must be adequately answered.

An attempt will be made to discuss and formulate the answer to above question, using mainly the results obtained with the aid of the HVS.

Average were calculated from three transverse and four longitudinal positions - a total of approximately 490 points.

0,20

0,/8

0,16

E

E

Z

0 j:

•...

..J

II.

•...

Q

0,14

0.12

0,'0

0,08

0,06

0,04

0,02

0,00

0

SECTION

:3

•••••

'0

20 30 40

!50

eo

70 80 90 100 110 120 130

'40

150

DISTANCE ALONG ROAD 'N METRES

0,22

0.20

Z

2

I-

•...

..J

II.

•...

Q o,'e

E

-

0.16

E 0,14

0.12

0,10

0,08

0,06

0,04 -

0,02

0,00

0

SECTION

~

1-..-J CONTROL

,.---, SECTION srCTION

I+4l

10 20 30 40 !l0

60 70 80 90 100 110 120 130 .40

1!l0

DISTANCE ALONG ROAD IN METRES

(b) OUTER NORTHBOUND

I AND 2 )

LANE (HVS TEST SECTIONS

FIGURE 3.2

BENKELMAN BEAM DEFLECTION OF THE ROAD

SURFACE AND POSITION OF THE DIFFEREN T

HVS TEST SECTIONS

It is believed that subbase layers are major structural elements in the pavement structure, because of their part in the reduction in surface deflection.

Experience has shown that almost all cementitious layers experienced cracking owing to traffic loading and changing environmental conditions.

Fairly early after construction shrinkage cracks develop owing to the loss of water as well as thermal conditions.

During the development of non-traffic associated cracking or thereafter, construction traffic is allowed.

Because of the dynamic action resulting from moving loads as well as subsurface water conditions, a gradual break down or "fracturing" of the cementitious layers occurred.

Some layers experience "active" and some layers "inactive" cracks.

The active crack is one which resulted in high relative horizontal and vertical movements

(Rust, 1985).

The inactive crack hardly experience any movements.

The activity of a crack depends also on its support and interlock which is not always constant.

Cracks on the surface often started off as inactive cracks and becomes very active with traffic loading depending on a range of factors such as :

(i) loading (rate and magnitude)

(ii) excessive porewater pressure and tyre pressure

(iii) cementitious characteristics of the cementitious layer

(iv) wet- and dry durability (erodibility)

(v)

(vi) support conditions underneath the cementitious layer temperature of asphalt

(vii)

(viii) compressive strength of the layers in the structure, etc.

fatigue characteristics of the cementitious layer and block sizes,

It should be possible with an HVS to quantify the gradual breakdown of the cementitious subbase layers.

It should also be possible to discover the effects of the initial reduction in vertical resilient deflection on subgrade layers.

The increase in resilient deflection during the

preand postcracked phases of the cementitious subbase layers can be monitored using multi depth deflectometers

(MOD).

The behaviour with an excess of porewater pressure in the postcracked phase of the structure should also be studied.

A typical layout of a test section is illustrated in Figure

3.3.

The average change in permanent deformation (rut development) on the three test sections at various moisture, temperature and loading conditions is illustrated in

Fi-

gure 3.4. The individual test progranunes for each section are summarized in Table 3.1.

As a standard for comparison it was decided to compare directly the change in permanent deformation on the three sections with the number of equivalent standard axles

(E80s).

A constant damage exponent of d = 4 was used in the

P d equivalency formula, (40) • The value of 4 does not neces-

sarily represent the "true" damage exponent for this structure.

The aim of these tests, however, was not to determine an accurate damage exponent, although some differences were noticed in the rate of change in the permanent deformation on the surface, using different wheel loads.

The permanent deformation of each section will be discussed separately.

LEGEND

T

M

B

ST

A

=

TEMPERATURE

=

MULTI DEPTH DEFLECTOMETER

=

BENKELMAN BEAM (RSDI

=

STRAIGHT EDGE

WATER INLET HOLES

"",

I,Om

I

B

B

B M B

B

I

?

I

,

B ,

B

I

B

B

,

I

I

I

B

8

,

Jf

B

.

_.

.

FIGURE 3.3

A TYPICAL LAYOUT

OF

AN HVS TEST SECTION

.

W

CD

f2

0 z

0 r=

C[

~

Q:

-

E 50

E

40

30

20

~

w

10

C> z

C[

:r: u

0

(a) j

50

I

SECTION t

2

ASPHALT TEMPERATURE: 20-30·C

LTSBs: LIME TREATED SUBBASES

100 kN DUAL-WHEEL LOAD

3

4

LOG(EBOs)

5

6

7

~ 40 t=

C[

:!

30

~

~ 20

~

10

~ z

C[

:r: u 012

( b ) SECTION 2

:!

lr

Q:

w

o

~ 50

Z o fi

40

30

20

3 4

LOG lE80s)

ASPHALT TEMPERATURE: x--«

30-45°C

.-..

20- 30°C

150 kN SINGLE-

WHEEL LOAD

WET

LTSBs

DRY

LTSBs

B

w

10

C> z

C[

5

0

(c

I 2

3 4

LOG lE80s)

5

I SECTION 3

FIGURE 3.4

AVERAGE CHANGE IN PERMANENT OEFORMATION

ON THE THREE TEST SECTIONS AT VARIOUS STAGES

OF TRAFFICKING

Section TrafficMoisture Aasphalt king load Condition Tempera-

Actual

Repeti-

Remarks

100

100

100

Dry

Dry

Wet

3 40

40

150

Dry

Dry

Dry

20 - 30

20 - 30

20 - 30

520 000 Total test section

175 000 Half of test section

175 000 Half of test section

695 000 Total repetitions applied

830 000 Overlapping of test sections

20 - 30

30 - 40

20 - 30

96 000 Half of test section

(Pts 9 - 15)*

96 000

Half of test section

(Pts 0 - 8)*

34 000 Total test section

130 000

Total repetitions applied

When considering the change in permanent deformation in Figure 3.4 on Section 1, which was tested with 100 kN dual wheel loading, the terminal level of 20 rom deformation was not reached during the test in the dry state (normal conditions) of the weakly cemented subbase layers, (LTSBs).

*

During the dry state, the rutting originated mainly within the asphalt base and surfacing layer.

An average rut of

13 mm was measured at the end of the test in the dry state.

The rate of change of rutting was calculated to be approximately 0,5 mm/MESO*.

During the wet test (i.e. the cracked subbases saturated), moisture-accelerated distress (MAD)

**

was experienced.

The origin of the rutting was seen to gradually change from the asphalt layers to the upper stabilized subbase layer, LTSB(l), in the wet state.

At the end of the test an average rut of 45 rom was measured on the wet section.

The asphalt surfacing and base layer were extensively cracked owing to fatigue.

Pumping occurred almost immediately after the cracks appeared on the surface of the test section.

(See photographic record of the tests in Appendix A.) The asphalt temperature varied between 20 and 30

DC which was the normal daily variation during the time of testing.

A total of 27 million E80s (MESOs) were applied to the dry test section, of which 6,8 MESOs were applied on the wet section.

After completion of the tests the stabilized subbase layers were extensively cracked. The upper subbase layer was more fractured than the lower subbase layer.

In the wet state, the former was granulated and the latter only fatigue cracked.

*

ME80s: million E80s

**

The terms in this report : wet, saturated and moisture accelerated distress (MAD)

1 all refers to a condition in the sublayers, especially between the asphalt and upper subbase layer, were excess porewater pressure developed during HVS testing.

The permanent deformation shown in Figure 3.4 is related to standard E80.

In Figure 3.5 the average deformation of the road surface at various stages of trafficking on Sections 1 and 2, is given against actual 100 kN repetitions.

Water was introduced from the surface to a depth of 440 mm on one half of the test section.

Deformations were measured with a

2 m straight edge.

A longitudinal permanent deformation profile of the surface of the section is shown in Figure 3.6.

The relative positions of the water inlet pipes used to introduce water into the structure are also shown in the figure.

Maximum deformations of 90 mm were recorded during this test after the introduction of water.

Subsequent cracking of the asphalt and pumping from the subbase layers were seen.

The deformation, however, was not uniform across the wet section.

Permanent deformations at different levels within the road structure were measured with the Multidepth deflectometers

(MODs).

Three MODs were installed on Section 1, as shown in previous Figure 3.3.

The average permanent deformation of the road surface at 'the three MOD positions at various stages of trafficking on Section 1 is illustrated in Figure 3.7(a).

A maximum of 82 mm was measured at MOD12.

Subsurface permanent deformations measured with the MODs at the three positions are shown in Figures 3.7(b), 3.7(c) and

3.7(d).

At MDD4 (Figure 3.7(b», the bottom of the asphalt base layer deformed approximately one millimeter, which was obviously the permanent deformation of the top of the upper stabilized subbase. The permanent deformation in the middle and bottom of the stabilized subbases were 0,3 mm and less, which is negligible.

In Figure 3.7(c) the permanent deformation in the sublayers measured with MOD8 is shown.

The upper MOD was situated 60 rom from the surface, within the asphalt base layer.

At this level deformations of up to

4 rom were measured.

The second MOD level was at 320 Mm, a location which is in the middle of the two subbase layers

(i.e. the top of lower subbase).

At this level the permanent deformation was found to be 0,6 rom and less.

Water was

50

LTSBs: LIME TREATED SUBBASES

.-.

SECTION 1 (DRY,WET)

45

E

E

40 z

Q

I-

<X

::i: a:

0 lL.

w

0

35

30

25

I-

Z

w

20 z

<X

::i: a:

w

Q..

15

10

5

0

WATER

(0-440mm)

TOTAL SECTION1

DRY LTSBs

/

."..

/pts 0-S(LTS8s-

DRY)

_._

.1

.-.;(.

.,-.,.,-x.-

•.....•. ......,...

--

._.·v....

./ .,..,-x--

.....-:._X

,.'~__--X--

X-- - X---X

200000 400000 600000 800000

NUMBER OF REPETITIONS (100 kN DUAL WHEEL LOAD)

FIGURE 3.5

AVERAGE PERMANENT DEFORMATION OF THE ROAD SURFACE AT

VARIOUS STAGES OF TRAFFICKING ON SECTIOIN f

AND

SECTION 2

8 METRE TEST SECTION

ACTUAL TEST SECTION

DRY SUBSURFACE

MDD4 t

4

I

MOD 8 t

I

WET SUBSURFACE

MOD 12 t

'¥ATER INLET PIPES o o

10

E 20

~ 30 z

40

2

50 le(

:E

60

Q:

70

~ 80

ILl a

90

X--X 10

0--0

200000

6--6 520000

0----0 678000}

---}

_

WATER INTRODUCED

_ 684 000 (0mm- 440 mml

X-X 695000

FIGURE 3.6

MAXIMUM LONGITUDINAL PERMANENT DEFORMATION ON THE SURFACE OF HVS

SECTION' AT VARIOUS STAGES OF TRAFFICKING

1IU11I£II CWII£I'[TITIOIIS 100 kll(_ lOll

-'00

100 700

100

ZOO

-'00 j

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~

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L£G£ND: fO_ zoo -'00 400 500 100

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100 700

100 100 -'00 400 500 100 i

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OEFORMAriON IN mE SUB -LArERS Ar

MEASURING POINTS 12

DEFORMATIONIN rHE SUB-LArERS Ar

MEASURING POINrS B

FIGURE 3.7

AVERAGE PERMANENT DEFORMATION MEASURED WITHIN THE TEST SECTION

WI TH THE MULTI- DEPTH DEFLECTOMETER (MOO) AT VARIOUS STAGES OF

TRAFFICKING

W

•...

U1

introduced into the sublayers after 520 000 repetitions.

This was done approximately one meter from MOD8 towards

MOD12.

The effect of the water was seen after approximately

678 000 repetitions, see previous Figure 3.6.

In Figure 3.7(d), the permanent deformation in the sublayers measured with MOD12 is shown.

During the dry state, the permanent deformation at the bottom of the asphalt and within the weakly cemented layers resulted in values of approximately 1,3 rom and 0,6 rom, respectively.

This compares fairly consistent with those measured at MOD8 and

MDD4.

During the wet state, however, dramatic increases in the permanent deformations were recorded at the bottom of the asphalt and within the weakly cemented subbase layers.

At the end of the test the magnitude of these deformations exceeded the measuring range of the MODs.

At approximately

650 000 repetitions, the deformations were 14 rom at the bottom of the asphalt (Le.

top of upper subbase) , and

3,5 rom in the middle of the subbase layers (L e.

top of lower subbase layer).

At the bottom of the lower subbase the deformation was approximately 0,6 rom.

This result indicated a decrease in thickness of the stabilized subbase layers, especially the upper subbase.

The subbase layer can only deform (loss in thickness) due to shear, densification or erosion.

It is however believed that shear and erosion contributes largely to the deformations after the stabilized layer experienced fatigue and "crushing" failure in the dry shear and erosion within the upper subbase layer resulted in subsequent excessive permanent deformations of the road layers is excessive pumping of fine subbase material through extensive fatigue cracking within the asphalt layer.

(See photographic record in Appendix A) • During this phase a state of excess of porewater pressure is believed to be the main reason for distress. These pressures equal or exceeds the tyre pressure during the test, which was in this case

approximately 700 kPa.

Such forces are very destructive and are the main reason for failures if erodible subbase material exists in structures containing thick asphalt or concrete bases.

Section 2 was tested only in the dry state, with the ambient asphalt temperature varying between 20°C and 30 °C.

The average change in deformation is illustrated in the previous

Figures 3.4(b) and 3.5.

The trafficking dualwheel load was

100 kN.

Relatively good agreement between Section 1 and

Section 2, with respect to change in asphalt deformation in the dry state was obtained.

A total of 32 MESOs were applied to Section 2 and an average rut of 12 mm was measured at the end of the test.

The comparable rut depth was

13 mm on Section 1.

The layout of Section 2 was similar to Section 1.

Overlapping in a longitudinal direction of the section was done twice.

The overlapping distance was approximately two metres.

The permanent deformation shown in Figure 3.4(b) was measured on the part of the section which received the maximum number of repetitions, i.e.

830 000, at 100 kN dualwheel load.

The permanent deformation measurements against actual repetitions are shown in Figure 3.5.

The rate of increase in asphalt deformation compared favourably with the rate of deformation measured on Section I, during the dry subbase conditions.

Test pits made across Section 2 after completion of HVS testing confirmed that early fatigue cracking of both stabilized subbase layers after approximately 94 000 repetitions with the 100 kN dualwheel occurred.

The upper subbase experienced advanced cracking, in a similar way to that found on Section 1.

Because asphalt deformation occurred on Section 1 and 2 with

100 kN wheel loads during the temperature range of 20°C to

30 °C and with the material in the dry state, it was decided to test a third section at a standard dualwheel load of

40 kN at elevated temperatures.

150 k N (SINGLE

WHEEl- LOAD I

ASPHALT TEMP.

20 - 30·C

_40

E

E

Z

35 o i=

~30 a: o tt;

25 o

~ 20

1&1

Co:)

Z ct

5

15

DOWNWARD

CHANGE IO,Ol

MEASURING POINTS I- 8 (30 - 40·CI

MEASURING POINTS 9-'5(20-30·C I

39 52 65 78

NUMBER OFREPETITIONS

91 104

(x 10

3)

FIGURE 3.

e

AVERAGE DEFORMATION OF THE ROAD SURFACE AT VARIOUS

STAGES OF TRAFFICKING AT INDICATED WHEEL LOAD AND

ASPHALT TEMPERATURE ON SECTION 3

This was done by introducing heat on one half of the test section using heaters. Two ranges of temperature were achieved on Section 3 namely 20°C to 30 °C and 30°C to 40 °C.

The hotter asphalttest continued up to 96 000 E80s.

(See Figures 3.4(c) and 3.8.) The average rut depths measured at

96 000 E80s were 2 and 10 mm for both the normal and hot temperature ranges.

These results indicated that if temperatures between 30 - 40°C were experienced for long periods of time together with relative heavy loading, excessive asphalt deformation will occur.

It is, however, believed that the higher temperature range together with heavy loading would only prevail for very short periods of time, and therefore excessive asphalt deformation

is

not expected during the service life of this structure.

An excessively high singlewheel load (150 kN) test followed the higher temperature test on Section 3.

The asphalt temperature during this latter test varied between 20 and

30°C.

The objective of the latter test was to study any permanent deformation (or distress) which may occur in the dry state of the weakly stabilized subbase layers.

After

34 000 actual repetitions (i.e. 6,7ME80s) , the average rut was 50 mm.

This rutting occurred in the short space of time throughout a single night.

In Figure 3.8, the average deformation of the road surface at various stages of trafficking is shown.

A difference

in

rut depth of approximately 10 nun was measured on the surface after the temperature test, at 96 000 E80s.

It is believed that the relatively young age of the asphalt surfacing caused the deformation,

(+

6 months untrafficked).

Excessive deformation occurred during the 150 kN test (50 nun rut) after

34 000 repetitions.

After discovering the origin of the rutting, it was observed that more than 80 per cent of the rutting originated within the asphalt surfacing and base layer.

This indicates that the support layers (stabilized

subbases) are adequate, but the asphalt layers are sensitive to overloading (i.e. wheels with high contact pressures, 1

450 kPa).

No fatigue cracks were found at the bottom of the asphalt base layer after the test.

The two stabilized subbases were cracked, but still provided adequate support.

The upper subbase, however, was more fractured than the lower stabilized subbase, a condition which was also observed on both the previous test sections.

Section 1 and 2 were trafficked with 100 kN dualwheel loads and Section 3 with a 40 kN dualwheel load up to 96 000 repetitions after which the wheel load was changed to 150 kN singlewheel load.

The average transverse profiles of the road surface at various stages of trafficking on the three sections is shown in Figure 3.9.

The results indicate a distinctive difference in profiles after trafficking with

100 kN dualwheel and 150 kN singlewheel load.

Although the behaviour of the stabilized subbases were virtually the same

(cracking and crushing), the asphalt surfacing and base layer appeared to be more sensitive to loading in terms of permanent deformation when the subbases remains dry.

It is believed that the higher bearing capacity of dry subbases contributes largely to this.

In Figure 3.10, a comparison of the average radii of curvature (Re) and road surface deflection (RSD) of the three test sections, is shown.

Similar results were seen on the three test sections in the dry state.

The curvature decreased from more than a 1 000 to 250 m, whilst the deflection increased from 0,18 to 0,6 rom.

In the wet state (Section 1) the curvature decreased from 230 to less than 100 m, whilst the deflection increased from 0,5 to more than

1,5 rom.

According to Freeme et al (1984) this road structure reached the flexible state after approximately 7 ME80s,

5

E

E

ILl v

~

II::

::::>

III

0

~

II::

-5

-10

-15

~

-20 l-

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0

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0

II::

II.

-25

-30

0

-35

-40

0

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• i~

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,--~-N.15299B:....

\

.....

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, ..

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.

" N"474427 ,

'-\

\ / '-'

N=829965

/\.1

.V'"

-20

0 200

400;

600 800 1000 1200

1400

1500

TRANSVERSE DISTANCEOVF;RTEST SECTION

1"",,1

1800

2000

20

E

E

ILl

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::::>

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0 cr lE

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-15

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-25

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,,/

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'\

\

\

\

\

N=IO

------

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N = 96189140 kNI

:

,

,

/

I

SECTION 31

WHEEL LOAD: 40kN AND

\

'\

'1-

\(..

·:.t

e

~

1150'40~

'

,

/

150kNISINGLE I, ~ •• lORY)

'

,

,,"\

(

"

'\

'\

"

200 400 600

800 1000 1200 1400

1600

TRANSVERSE DISTANCEOVER TEST SECTIONImml

1800 2000

FIGURE 3.9

AVERAGE TRANSVERSE PROFILE OF THE ROAD

SURFACE AT VARIOUS STAGES OF TRAFFICKING

ON THE THREE SECTIONS

1400

0 e:[

Q:

ILl

Q:

•...

e:[

>

Q:

::l

U l&.

0 u

-

E 1200

Q:~

1000

600

400

200

0

E

~ 1500 o

If)

Q:~ z o

1250

~ 1000

It

750 o

ILl

U

~

Q:

500

;;: 25U

o e:[ o

Q:

0

-

- --

SECTION

SECTION 2

1

_.

SECTION 3

-

---

_.-

SECTION 1

SECTION 2

SECT ION 3

WATER INTRODUCED

INTO SECTION I

2 3 4

LOG (EBOI

FIGURE 3.10

COMPARISON OF THE AVERAGE RADII OF CURVA-

TURE AND ROAD SURFACE DEFLECTION

MEASURED ON THE THREE TEST SECTIONS AT

MARIANNHILL, N3/1

flexible state is possible only during saturated conditions especially of the upper weakly cemented subbase, when fractured (RSD>0,6 mm).

Owing to increase in cracking and fracturing of the upper stabilized subbase, the layer's sensitivity to water, i.e. breakdown, softening and potential for pumping (erodibility) increased.

In Figure 3.ll(a), (b) and (c) average longitudinal road surface deflections at various stages of trafficking on the three sectors are shown.

In Figure 3.ll(c), the deflections under a 40 kN dual wheel load on Section 1 are illustrated.

The results indicate the effect of the excess porewater pressure within the subbases of this section, reflected by the rapid increase in deflections from approximately 651 000 repetitions in the wet state.

Lower deflections resulted on the dry part of the section.

The same irregularity the deflection measurements existed as with the permanent deformation resulted on this section.

See previous Figure

3.6.

This

is

an indication of variability in road behasubbase material could decrease the risk rapid increase in

RSD and subsequent failure.

In Figure 3.11 (b) the deflections resulted on Section 2 are shown.

This section was trafficked

in

the dry state only.

After approximately 830

000 repetitions with the 100 kN wheel load the maximum deflections were approximately 0,55 Mm.

If a damage exponent of 4 is used to convert these repetitions to E80s, almost 33 x 106 E80s had been applied to this section without cracking or serious deformation.

This indicates the relatively high structural capacity of this type of pavement in the dry state.

It is believed that the weakly cemented subbase layers are the main structural elements in this structure, causing the relatively low surface deflections, even after 30 x 106 E80s.

The asphalt mixes also proved to withstand excessive permanent deformation during normal trafficking and prevailing temperature.

3,0

2,e

2,6

214

IL

I&J

0

-

2,2

-

2,0

E e

I,e z

1,6

0

~

I," u

I&J

..J

1,2

1,0 o,e

0,6

0,4

0,2

0

MDDI2 MEASURING LOAD I 40kN tRAFFICKING !DAD: 100 kN

MOD"

MoDe

I

+

N651175lWETI

.,....

:2

3 4 5 6

7 e 9 10

N52000010/tY/

••

l<

X

..,

"""'V'"--"

X

--.<

12 13 14 15

3,0

2,8

2,6

2,"

2,2

E

E

-

2,0

I,e z

1,6

~

1,4.

~ u

I&J

..J

1,2

IL

I&J

1,0

Q o,e

0,6

0,4

0,2

0 123"

~'Et.SURING LOAD !"OkPl

TRAFFICKING LOAD: 100 kN

5 6789

LEGEND

)Io-oof(

NIO

•••••• N60!lS6

.-.

N92"42 b--4

N 152998

~ N559800

*-*

N829965

10111213 1415161711119202122232"~2627282930

MEASURING POINTS (x 500 mm I

3,0

2,8

2,6

2,4

E

2,2

-

2,0

E 1,8

Z

~

~

U

I&J

..J

IL

I&J

Q

1,6

1,4

1,2

1,0 -

0,8

0,6

0,4

0,2

MEA.SURING

LOAD: 40kN,I50kN

TRAFFICKING LOAD: 40 kN (OUALl 0-961119 REPETITIONS

150kN (SINGLE) 9618.

-129138 REPETI-

TIONS

MDD4

~"'O'NI

MODe t

MDDI2

'

I

0

(C )

2 3 4 5 6 7 8 9 10 "

MEASURING POINTS ex 500mml

SECTION! (40kN,150kN)

12 13 14 15

FIGURE 3.11

AVERAGE LONGITUDINAL ROAD SURFACE DEFLECTION

UNDER THE INDICATED WHEEL LOADS AT VARIOUS

STAGES OF TRAFFICKING ON THE THREE SECTIONS

In Figure 3.11(c), the deflections resulted on Section 3 are illustrated.

At the end of this test the 40 kN deflections were less than 0,66 rom,which compared favourable with those measured on the previous sections.

The surface deflection under alSO kN single wheel load varied between 1,0 mm and

1,6 rom.

These deflections are indicative of the sensitivity of this pavement structure to heavy overloading when the weakly cemented subbase layers both in the precracked and postcracked phases when dry.

In order to evaluate the elastic response of the pavement structure, it was decided to measure the surface deflection at various wheel loads. Loads from 20 kN to 100 kN were used on Section 1 and 2 and single wheel loads from 20 kN to

200 kN were used on Section 3.

The average road surface deflection and radii of curvature (RC) of the three sections are illustrated in Figure 3.12 (a), (b) and (c).

All the sections behaved linear elastic for wheel loads up to 70 kN,

Le.

the deflection doubles as the wheel load doubles.

Initially (N

=

10), wheel loads higher than 70 kN resulted in non-linear elastic responses (softening) from the sections.

This "softening" however changed to "stiffening" after the HVS trafficking, which is also a non-linear elastic response.

The RC results appears to be more linear with increase in trafficking.

The linearity of RC however appears not to be an indication of linear elastic response of the pavement structure.

Relatively high RC's were measured under a 40 kN wheel load, at the end of the HVS test (250 m 500 m).

In Figure 3.12 (a), the resilient response of the control section is also indicated and is consistent with the results measured on the other sections.

It is again believed that the main structural element in this structure, to reflect such linear elastic behaviour is the two weakly cemented subbase layers, even in a cracked state (dry).

~

~

II:

E

;;:; 2000

II:

:;)

I-

;!

1~OO -

II:

:;)

...

1000 o

~OO

0

I,~ __

------

• LINEAR ELASTIC RESPONSE

CONTROL SECTION

0,5

....--

.

....__

o o

0)

20

SECTION 1

••

40

LOADI kNI

60

~

..

~; tl' \0

~

II>

02

Q

<l

II:

; 1500

~ a

1000

500

E

!

z

S!

I-

U

...

•••

Q

II>

:;) i5

<l

II:

E

;;:; 2000

~ o o

20 b) SECTION 2

~ I~OO

II: a

~ 1000

500

0

1,5

1,0

0,5

0

0 c)

20 40

SECTION

3

60

40

LOADlkN)

80 100

LOAD (kN I

120 140 160 180 200

FIGURE 3.12

AVERAGE ROAD SURFACE DEFLECTION AND RADII

OF CURVATURE UNDER VARIOUSDUAL WHEEL

LOADS ON THE THREE TEST SECTIONS AT

DIFFERENT STAGES OF TRAFFICKING

Resilient depth deflections were measured with the aid of

MOD's.

The average MOD depth def:'ection measurements at various stages of trafficking on the three test sections under a 40 kN dualwheel measuring load are illustrated in

Figure 3.13.

The figure indicates relatively small relative deflections resulting from both weakly cemented subbase layers.

The majority of surface deflections are attributable to the selected subgrade layer downwards.

Higher relative deflections were also measured within the asphalt layers.

During the wet state (Section 1 only), the relative deflection in all the layers, subbase included, increased markedly.

(See

EPwp· state (MOD12) in the figure).

The resilient structural capacity of the cemented layers is reflected by the comparatively low relative deflections.

According to De Beer, 1984 (a) the behaviour of all the layers in this structure is linear elastic (axle loads up to

80 kN) • It is therefore concluded that linear elastic analysis of this pavement structure would probably be better than non-linear analyses.

According to beam flexural strength tests, visible cracks in the stabilized subbase material appeared at deflections ranging between 0,1 rom and 0,3 rom (De Beer,

1984 (a» • From this information it is quite possible that when these levels of deflection were reached during the HVS tests, cracking occurred within the stabilized layers. It is therefore possible that cracking of the weakly cemented subbase layers occurred when the deflections on top of the selected subgrade increased (rapidly) beyond the 70 kN wheel load.

Test pits made after HVS trafficking revealed some crack patterns printed on top of the selected subgrade layer which can only be the result of cracked subbase layers.

SURFACING

BASE

0

100

200

LTSB

300

LTSB

400

!l00

SSG

600

700

E

E

800

:r:

•.. 1000 lL

III

Q

SG

-

---

DEFLECTioN 1",,,,1

1,(1

-

I, 1,4

1,6

--

100

200

0

300

/

400

!I 00

600

700

800

ISECTION

Il

DRY

EPWP IMDOl21

'TRAFFICKING WHEEL LOAD,IOOIlN

1000

2000

I,D 0

DEFLECTION t",,,,1

02 o

10.

27,B .10·

32,3.10·

}

E80.

I

SECTION

DRY LTSeS

21

100

200

300

400

I

!l00

600

700

800

1000

0

0,'.10·

I

I

I

~O.IO·

I

'r.~bf-

-7----

,

I

I

I

SECTION

31

DRY L TSeS

I

I

I

TRAFFICKING WHEELLOAD:

TRAFFICKING WHEEL

LOAO: 100llN

2000 1111

2000

FIGURE 3.13

AVERAGE MDD MEASUREMENTS AT VARIOUS STAGES OF TRAFFICKING ON THE

THREE TEST SECTIONS UNDER A 40kN DUAL - WHEEL LOAD

(X)

.

w

N

The deflection on the top of the weakly cemented subbase layers increased from approximately 0,2 mm to 0,58 mm during the dry state on Section 1.

During the EPWF state the deflection increased to approximately 0,95 Mm.

At the bottom of the lower cemented layer (top of selected subgrade layer) the deflection increased to approximately 0,7 mm which can be compared to the initial 0,8 mm RSD measured on the selected subgrade layer after construction.

This is an indication of the enormous reduction (almost 100 per cent) in load carrying capacity of the upper layers during the

EPWF state.

The surface deflection under the 40 kN dual wheel load increased from 0,15 mm to almost 1,6 mm during the dry and wet states at MOD 12.

The higher resilient deflections is an indication of the loss in the structural capacity owing to excess porewater pressure which existed not only on top of the upper subbase layer, but also within the layer.

Because the layer developed an advanced state of cracking in the dry state, (confirmed by cracked recovered subbase material after HVS tests), voids existed which were then filled (saturated) with water.

With the trafficking wheel load (overburden pressure) the EPWF state develops.

Enormous pressures are created thereby "blowing the upper pavement layers into pieces"!

In this state (Section 1), the support from the subbase layers decreased, with the asphalt layers consequently becoming overstrained.

This resulted in excessive fatigue cracking of the asphalt, pumping and finally excessive rutting on the surface.

See photographic record of HVS tests in Appendix A.

Mechanistic analyses were done on the structure, using the

RSD and MOD results.

The linear elastic analysis was done with the ELSYM 3 programme together with failure criteria proposed by Freeme et al (1984).

The MOD deflections were used to calculate effective elastic moduli values of the

different layers.

The moduli values and assumed Poisson's ratio values were then used to calculate stresses and strains in the simulated model.

Moduli values, measured and calculated deflections for the different test sections and different stages of trafficking are summed in Tables 3.2, 3.3

and 3.4.

Good agreement exists between the measured and calculated deflections.

The initial effective stiffness of the asphalt surfacing and base layer varied between 4 000 and 1 500 MFa.

In the dry state, the effective stiffness decreased to a range of 1 250 and 570 MFa.

In the wet state, however, the value decreased almost to 12 MFa.

The asphalt layer experienced fatigue cracking and stripped during the wet test.

The modulus of the upper stabilized subbase, varied initially between 9 900 and 7 400 MFa.

After 25 ME80s the value decreased to a range of 1 000 to 800 MFa.

In the wet state the value decreased to 260 MFa.

The effective elastic modulus of the lower stabilized subbase, varied initially between 3 600 MPa and 3 000 MPa.

The modulus decreased to a range of 700 to 400 MPa after 25 ME80s.

In the wet state, the modulus decreased to 200 MFa.

During the high singlewheel load test (150 kN) in the dry state, the modulus decreased also to almost 200 MFa.

The modulus of the subgrade layers ranged initially between

110 and 450 MPa.

This value decreased to a range of 40 to

60 MPa after 25 ME80s.

The modulus, however, did not decrease during the wet state, but increased from 41 to

50 MPa, showing a slight "stress stiffening" behaviour probably because the subbase layers lost its load bearing capacity resulting in higher stresses to be transferred to the subgrade layers.

TABLE 3.2 - The calculated effective elastic moduli values together with the measured and calculated depth deflections at various stages of trafficking on Section

1.

HVS TRAFFIC (ME80s) DEPTH LAYER

(mm)

0 4 8,3 25,4 26,5

26,5

(semi-wet-MOD8) (wet-MODl2)

Deflection

(pm)

Mea-

Calcusured lated

E

Deflection

(MPa)

(pm)

Mea-

Calcusured lated

E

Deflection

(MPa)

\.tm)

Measured

Calculated

E Deflection E

(MFa)

\.tm)

(MFa)

Deflection

(Pll)

E Deflection

(MPa) (pm)

MeaCalcu-

Mea-

CalcuMea-

Calcusured lated sured lated sured lated

E

(MFa)

800 500 493 790 600 594 750 746 734 500 1 670 1 612 12 0 Surface

& Base

180 177 2 799 350 346

165

LTSB(l)

160 159 7 428 280 279 2 200 430 430 2 200 530 530 800

637 636

550

890 886 260

315

LTSB(2)

155 155 3 599 265 267 2 058 415 416 1190 490 496 672 584 585 350

760 757 200

465 selected

150 subgrade

150 110 255 257 74 400

401 44 465 471 41 543 543 40 660 666 50

.

w

The tensile and compressive strain values, together with the appropriate failure criteria, were used to predict the life expectancies (structural capacities) for the different layers in the structure.

In Table 3.5

a summary of the different strain parameters (distress determinants) used, are given.

In Table 3.6

the strain values for the different sections at various stages of testing are given.

Similar strain values were obtained on Section 2 and 3.

The strains increased rapidly during the EPWP state on Section 1.

The predicted E80s (using the current failure criteria) to result in visible cracking on the surface of the layers and limit subgrade deformation for the three sections are summarized in Table 3.7.

The values were derived using the mechanistic design method proposed by Freeme et al (1984).

The HVS results indicated that the strains initiated at the bottom of the relatively stiff (high moduli) layers, i.e.

asphalt and cementitious layers, are not constant.

It varies with the number of load repetitions.

The rate of increase in these strain values are a strong function of the moisture and crack state of the layer itself and the support from the other layers in the structure.

TABLE 3.5 - Summary of different strain parameters used in the analysis

Horizontal tensile,

E t Bottom of layer

(crack initiation)

Upper stabilized subbase

Lower stabilized subbase

Selected and subgrade layer

Vertical compressive,

E

Top of selected v layer (Umi t permanent deformation. )

TABLE 3.6 - A summary of the calculated strain values at various stages of trafficking in the different layers of all three test sections

ME80s

(HVS)

8,3

25,4

26,5(SW)*

26,5(W)

Bottom of

Asphalt

£t (\1£)

8

38

30

135

203

360

LTSB(l)

£t(\I£)

21

36

63

102

179

331

LTSB(2)

£t(\I£)

36

74

108

165

242

421

(b) SECTION

2

ME80s

(HVS)

0

·3,6

7,5

27,8

32,3

Bottom of

Asphalt

£t(\I£)

0,4

Nil

8

83

156

LTSB(l)

£t(\I£)

21

7

28

94

146

LTSB(2)

£t(\I£)

33

33

63

115

134

Top of

Selected Layer

£v(\I£)

-39

-82

-119

-188

-293

-596

Top of

Selected Layer

£ (\1£) v

-39

-35

-67

-139

-284

ME80s

(HVS)

Bottom of

0,1

>50

Asphalt

£t (\1£)

12

4

132

LTSB(l)

£t(\I£)

18

10

173

LTSB(2)

£t (\1£)

23

25

174

* SW

W

Semi-wet: Subba8es wet but not saturated

Wet: Subbases saturated (excess porewater pressure state)

Top of

Selected Layer

£ (1I£) v

-29

-27

-244

TABLE 3.7

- Predicted E80s to cracking and limit subqrade deformation on the three test sections

Applied

E80a

(HVS)

(ME80s)

Asphalt LTSB(l)

0

4

8,3

25,4

26,5

(SW)

26,5

(W)

>50 x 106

>50 x 106

>50 x 106

>50 x 106

8 x 106

600 000

>50 x 106

>50 x 106

1,8 x 106

600 000

0

0

(SW) See Table 3.8

(W) see Table

3.8

(b)

SECTION 2

Applied

E80.

(BVS)

Cracking on the surface of

3,6

7,5

27,8

32,3

Asphalt

>50 x 106

>50 x 106

>50 x 106

20 x 106

8 x 106

LTSB(l)

>50 x 106

>50 x 106

>50 x 106

21 000

0

LTSB(2)

9 x 106.

300 000

0

0

3 000

0

>50 x 106

>50 x 106

>50 x 106

>50 x 106

>50 x 106

100 000

Limit top of subqrade selected

LTSB(2)

21 x 106

21 x 106

1,2 x 106

120 000

27 000

>50 x 106

>50 x 106

>50 x 106

>50 x 106

>50 x 106 strain layer at the

Applied

E80.

(HVS)

Cracking on the surface of

Asphalt

>50 x 106

>50 x 106

3 x 106

LTSB(l)

>50 x 106

>50 x 106

0

Limit subqrade strain at the top of selected layer

LTSB(2)

>50 x 106

>50 x 106

0

>50 x 106

>50 x 106

>50 x 106

According to Table 3.7, after 25 ME80s, the strain level

(effective horizontal tensile strain,

(E: t

» in the asphalt was relatively low and the structural capacity was calculated to be in excess of 50 ~~80s.

After the introduction of water, the fatigue life reduced to approximately 600 000

E80s.

Surface cracks on the asphalt occurred however just after the introduction of water into the subbase layers.

According to the prediction of cracking for weakly cemented subbase layers, the lower subbase layer should develop fatigue cracking long before the upper subbase layer.

The

HVS tests proved that the real situation is visa versa.

The upper subbase was in an advanced state of fatigue cracking, than that observed in the lower subbase layer.

Field inspections revealed that the interface between weakly cemented layers was relatively smooth and of utmost importance.

If a smooth surface exists between these layers, the two subbases does not react as a solid beam (thickness

300 mm), but reacts as two separate beams (2 x 150 mm) placed on top of one other with relatively low horizontal friction at their interface.

It can be shown mechanistically that relatively high horizontal strains will be experienced at the bottom of the upper layer by inserting a thin zero modulus layer between the two subbases.

(See

Chapter 5 later). The advanced state of fatigue cracking in the upper layer can then be explained.

The main disadvantage in the current mechanistic design criteria for fatigue life of treated layers lies in the fact that strain values at the bottom of treated layers change with trafficking.

Because of this behaviour, it is meaningless to predict structural capacities for the different layers, especially cementitious layers, using the current procedure, unless the real strain value at that time is known.

According to the prediction on Section 2, (See Table 3.7) surface cracking on the asphalt layer will occur after approximately 40 ME80s

(32

+ a MEaOs), provided the horizontal strain at the bottom of the asphalt layer being kept at a value of approximately

160

~E:

(Table 3.6).

The same reasoning holds for the weakly cemented subbase layer on Section 2.

No cracking is expec-

ted in the upper laYer after 8 ME80s, provided the strain level being kept constant at 28

~£.

This is possibly correct, but the practice indicates that the strain value is not a constant.

Therefore the limitations of the current design and evaluation criteria are realized.

A "performance expectation chart" must be constructed to enable future predictions rather than the use of the current "design curves" , based on actual induced strain vs repetitions history.

The deformation criteria proposed by Freeme et. al. 1984 for the subgrade layers in this type of design proved to be adequate.

The changes in the vertical strains are relatively small and therefore no real deformation problems for this layers are foreseen.

According to Tables 3.2 and 3.3, the vertical deflection at the bottom of the lower subbase varied initially from

0,09 mm to 0,50 mm during the dry state.

The relative deflection within the two layers varied between 0,04 mm and

0,10 Mm.

This is accompanied by effective elastic modulus changes of

>

5 000 to less than 500 MPa in the dry state.

In the EPWP state the relative deflection increased to

0,9 mm and the modulus decreased to 200 MPa, on Section 1.

(The relatively low moduli values calculated for the dry subbase layers on Section 2 (Table 3.3) after 32,2 ME80s are believed to be due to malfunctioning of the MOD's.) The moduli values for the subbase layers after 27,8 ME80s were more reasonable varying between 400 to 1 000 MPa in the dry state.

In terms of moduli it is concluded that the weakly cemented subbases still provided good structural support at the end of the test in the dry state.

This was also found with a high standard crushed stone base pavement in Transvaal, Van Zyl (1983).

Inspection trenches made after HVS testing enabled densities and moisture contents to be measured.

Comparisons of trafficked vs untrafficked sections were also done.

The position of these trenches on Section 1 are illustrated in

Figure 3.14.

In Figure 3.15 cross section profiles at the two trenches on Section 1, are illustrated.

A distinct difference was observed between the dry (wall A) and wet

(wall D) sections.

The origin of the surface permanent deformation (rut) during the wet state (EPWP) was within the granulated upper weakly cemented subbase.

During the dry state only fatigue distress occurred within both weakly cemented subbase layers.

It was not possible to measure any decrease in density of the weakly cemented subbase layers during the dry state of testing, however, increases of approximately 1,5 to 3,2 per cent were measured in the upper and lower subbase layers, respectively.

The insitu moisture contents varied between 9,6 per cent for the upper subbase and 13,8 per cent of the lower subbase layer.

During the wet state however a decrease of 4,3 per cent in density of the upper subbase was measured.

In the lower subbase (not granulated) a slight increase of 0,6 per cent was measured.

The average moisture content of the upper subbase was

9,2 per cent.

The reason for this relatively low moisture content

is

that moisture samples were taken approximately two weeks after the end of the test, therefore natural drainage resulted in moisture contents more or less equal to the moisture contents in the dry state at the time of sampling.

An increase of 4,8 per cent in density was observed within the selected layer.

It is believed that the decrease in density (strength) of the upper subbase, causes increased load transfer to the selected layer, which compacted under traffic resulting in relatively higher densities.

X

OCP,

• MOD

*

MAXIMUM RUT - 80mm

WALL 0 (WET)

WALL B (SLIGHTLY MOIST)

WALL C (MOIST)

WALL A (DRY)

DIRECTION OF

.HVS TRAFFIC

X

FIGURE 3.14

POSITIONS OF THE TWO TRENCHES EXCAVATEOON TEST SECTION f

LONGITUDINAL SECTION

(m)

0,2 0,4 0,6 0.8

1,0 I 2

BASE

UPPER

SUBBASE

LOWER

SUBBASE

SELECTED

SUB GRADE

LONGITUDINAL SECTION ( m)

00,0 0.2

0,4 0.6

0.8

1.0

1,2

I.e

BASE

100

-200

E

300

;; 400 /" / / / " / / / / // /;" / / t

500

'0':'

·0:-:'0"'.0

.:.0 ..

0 ...

0.... 0... 0.

1IJ

600 .'

'.'0-:":- ..

0:

o

700 ....

0 ".

800

.0 . ..

".

:d'

'-0'

"1

:"0" :.

.' .' o· .. · ..

0:.....

0::'

0 . : ... :

....:6 ....

TEST SECTION ••

:0'· :-'.. :

o· .. ' .' '. O· .

0 '..'.

·0·.·..

\" ..

:'?<'~

:0

,,o ..

0'.

UPPER~

SUBBASE

LOWER

SUBBASE

SELECTED

SUBGRADE

SUBGRADE

00,0 0,2 0,4 0,6 0,8 1,0 1,2 1,4

100 BASE

200

300

-

~ 400 /////

·:·q~·~5~l.

SECTION

.

'

.. .

UPPER

SUBBASE

500 ,'.' ..

l)' " .' .

:I:

600

:0·. ,

1i::

0·' ."

'0 ' (:)'. ' ..'..'

: ,"0.'· : "..

·.·.0

1IJ

700 . ,: .. 0 ... o

800 '.

·0.'.·:.·

0....

. ' " o.

,...

0 .. '.'

· ...0., .'

0 ,

',0' ...~.Q

'.

.0, .

'0': .:.

'0:

.', '.' ..

900

/~ /// //

.

"

///

LOWER

SUBBASE

SELECTED

SUBGRADE

~1~c?,·:.~."P"

,0:0·'

."

.

00,0 0,2 0,4 0,6 0,8 1,0 1,2 1,4 1,6

BASE

100

200

300

E 400 ///", /

5.

ti:

'.

500 '0"'0,"0'··

.' .

'., '.' .

600 ~

"

q :.:.'.'.

'? . :

0'·.

'0': :-

/

0 : .. " '. .....

0:'

0

////

/"

"'0'

/

:'.0.: :

W

0

700 .' ':'

800

.·0· ..

·.0:·

0':· ....

" ' ..

, ..

....'

0 '.

··b·:·: l>'~

.

.

0'. :::"::

0.:

900 .)::( ~:.:,

:C! '. '~ .••

r-

TEST SECT ION

"1:.O~·.:.o.::;:

':0'.

q-:

UPPER

SUBBASE

LOWER

SUBBASE

SELECTED

SUBGRADE

SUBGRADE

WALLO (WET)

FIGURE 3.15

CROSS SECTION PROFILES AT THE TWO TRENCHES ON SECTION f

.

w

I.

On Sections 2 and 3 similar densities and moisture contents were measured as in the dry state on Section 1.

It is interesting to note that during the high single wheelload

(150 ~~) test on Section 3 the density of the asphalt increased by approximately 5 per cent.

It is believed that the relatively stiff weakly cemented subbase layers provided strong support to the load therefore resulting in the densification of the asphalt layers.

No decrease in density was measured within both subbase layers although these layers were fatigue cracked into small blocks (approximately 150 nun x 150 nun) •

The density in the selected and subgrade layers increased by 2,4 and 3,2 per cent, respectively.

According to the mod. AASHTO values, the density increased from 94 per cent to 96 per cent in the selected layer and 98 per cent to 101 per cent in the subgrade layer.

On Section 3 (150 kN test), 1 to 2 per cent lower moisture content values were measured within the selected and subgrade layers and were associated with the more dense areas

(trafficked areas).

A cross profile made on Section 3 is illustrated in Figure 3.16.

The profile was made at the position of maximum rut (60 nun) on the section.

More than

80 per cent of the permanent deformation on the surface occurred within the asphalt surfacing and base layer only.

It was however not possible to distinguish accurately between the surfacing and the base layer.

Between 10 and

20 mm permanent deformation occurred on the surface of the upper cemented layer, and a maximum of 10 rom occurred within the selected subgrade layer.

The two subbase layers were cracked, with the upper subbase more fractured than the lower subbase layer.

The lower layer experienced only fatigue cracking.

In order to supplement the density results on the weakly cemented subbase layers, it was decided to measure water permeability on the two subbase layers, using the Marvi1 apparatus (Vi1joen and Van Zy1, 1983).

This was done in

E

!

300

:; 400 o l~

WIOTH Of TEST ~

SECTION -

I

CROSS-SECTION

0,2 0,4 0,6 0,8 IP

Imm I

',2 ~4 1,11 ~.

FIGURE 3.16

CROSS PROFILE OF SECTION ~ AT THE POSITION OF

MAXIMUM RUT

situ (on the trafficked material) as well as on recovered blocks of weakly cemented material after trafficking on

Section 2 only.

The in situ permeabilities were done from the top of the layers.

The permeabilities on the recovered blocks were done from the bottom of the layers.

The recovered block of material was turned upside down before the measurement.

The permeabilities were done at the same temperature and with the same source of water.

Road surface temperature ranging between 20 - 30°C (on the asphalt) at the time of testing.

Permeabilities were also done on the asphalt layer but this layer proved to be relatively impermeable using the MARVIL instrument.

The permeability results are illustrated in Figure 3.17.

In Figure 3.17(a), the permeability measured on the upper subbase are illustrated.

Relatively higher permeabilities were obtained from the bottom of the recovered blocks of subbase material.

The blocks however were handled very carefully and the higher rate is believed to be a result of invisible hair cracks

(fatigue) at the bottom of the weakly cemented layers.

The figure also indicates relatively higher permeabilities on the blocks where the higher number of repetitions were applied, although there is not clear evidence that increases in permeability occurred with number of repetitions.

In

Figure 3.17 (b) the permeabili ties measured on the lower subbase at various stages of trafficking are illustrated.

Relatively higher permeabilities also resulted on the recovered blocks of subbase material.

Relatively lower permeabili ties, however, were measured in the lower subbase.

This confirms the relatively higher degree of cracking observed within the upper subbase.

The in situ dry densities and nuclear moisture content values are also indicated on the figure.

350

300

250

9

I&,.

~

..,

E 200

150

100

50

LEGEND:

••••••

REPETITIONS

92442

If-ooK 413000

D-C 416000 b--t," 736000

0--0 829000

ON RECOVERED BL.OCK

--IN SITU

I )

DRY DENSITY

-------

I:

J

MOISTURE CONTENT

-_

•........

-

...

-

....

-

10

TIME

I min.)

15

FIGURE 3.17

MARVIL PERMEABILITIES MEASURED ON THE STABILIZED

SUBBASES ON SECTION

2

A T VARIOUS STAGES OF TRAFFICKING

Dynamic cone penetrometer (DCP) tests were done on the dry and wet parts of Section 1.

The DCPs were done from the top of the upper subbase.

The main objective with this tests was to quantify the state of the individual pavement layers in terms of DCP value (i.e. shear strength).

The average penetration rate was approximately 0,8 nun per blow in the upper subbase in the dry state.

The penetration rate outside the trafficked section was approximately 0,5 mm per blow.

See Table 3.8.

These are very low penetration rates

(Le.

higher strengths) but the higher rate of 0,8 mm per blow within the test section is an indication of weakening of the upper subbase owing to traffic. The average penetration rates within the lower subbase, selected subgrade and subgrade, however, were 1,5; 7,5 and 2,5 mm per blow, respectively.

*

On the wet section, the penetration rate of 2,2 is the same than those measured crushed stone base pavements in Transvaal (Maree, 1982).

The results indicates that traffic loading causes break up or cracking of the stabilized layers (loss in shear strength).

The average block size (diameter) measured from the upper subbase material on the dry section was approximately 250 mIn.

On the wet section the upper subbase was granulated and the particle sizes average at nominal diameter of approximately 50 mIn. Horizontal cracking was also observed within the upper subbase layer, after trafficking.

*It is important to note that the DCPs were done in order to quantify the state of the subbase layers rather than to quantify the balance of the structure.

Therefore the balance curves proposed by

K1eyn et al (1983) are not included in this study.

In Table 3.8 a summary of DCP penetration rates observed on the three test sections, is given.

TABLE 3.8 - Averaqe DCP penetration rates for the three Sections in

DIl per blow

Section

3

1

1

2

Subbase layer Position* Number of repetitions

Wheel load (1eN)

Inside

Upper

(dry)

Lower (dryl

Upper (wetl

Lower

(wet)

Upper

(dryl w,;'er (dry)

Upper

(dry)

Lower (dry)

0,8

1,5

2,2

1,5

0,6

2,0

1,1

2,0

Outside

0,5

0,5

<0,5

0,6

0,5

1,1

0,5

2,0

695 000

695 000

520 000(drylJ175 ooO(wetl

520 000(drylJ175 OOO(wet)

830 OOO(dry)

83 ooO(dry)

96 000 (4C1tN)J34 000(150kN)

96 000 (40kN) J34 000 (150kN)

100

100

100

100

100

100

40,150

40,150

* Relative to HVS test section

From the table it can be concluded that although the penetration rates were relatively low, the highest penetration rate of 2,2 rom per blow was measured in the upper subbase after the wet test on Section 1.

During trafficking in the dry state, penetration rates increased between 20 to 180 per cent in the upper subbase and 30 to 400 per cent in the content of the lower subbase is the major factor contributing to the higher percentage increase in penetration rates, in that moisture content is a critical parameter in the shear strength of soil materials.

As no marked differences in the moisture contents within and without the sections were obtained, the increase in penetration rates in the upper subbase is believed to be mainly due to trafficking.

The majority of the HVS results indicated that cracking

(postcracked phase) of cementitious materials is a reality.

(See also photographic record of HVS tests in Appendix A).

Because of the geometry of the layers in the structure and the loading, the material initially break up or crack owing to fatigue (beam bending concept).

At the stage, when the average block size (diameter) reaches the layer thickness, t, the probability of further fatigue distress is almost zero.

The only mode of further break up is therefore fracturing, crushing and/or shear, depending also on the porewater pressure in or on the material.

In

Table 3.9 an indication of different block sizes is given for the upper subbase layer at various stages of trafficking.

of trafficking on the structure

Number of repetitions*

100 kN 40 kN (E80s)

State of cracking

Estimated Block

Size

(diameter)

"Dry

EPWP**

0

92 000

520 000

830 000

0

3,6 x 106

20 x 106

32 x 106

Precracked

Postcracked

Postcracked

Postcracked

Slab

4 t

3 t

1,5 t

* 40 kN repetitions (E80s) calculated with d t

=

4, in

(L)

d

40

= layer thickness (t

=

150 mm in this case)

Slab

4 t

2 t

<0,3 t (Granulated)

On Section 1 it was observed that after block sizes reached dimensions of approximately 1,5 t - 2 t in the dry state, only 34 000 repetitions at 100 kN wheel load (1,4 x 106

E80s) is needed in the EPWP state to change the material into a granulated state (block size ~ 0,3 t).

The lower subbase only break down to sizes of the layer thickness even in the EPWP state because the stress level is markedly reduced at this depth in the pavement structure.

It is interesting to note that Otte (1978) also found that cementitious base layers do crack and that the average block sizes reached equals approximately the layer thickness.

In this case the base layer was 100 mm thick and blocks of 100 mm x

100 mm were recovered after HVS trafficking.

In order to summarize this pavement

I s behaviour and to indicate the important function of the weakly cemented subbase layers (especially the upper subbase), Figure 3.18

was compiled.

The figure indicates resilient and permanent changes of the pavement structure·with increase in traffic loading and change in moisture condition.

The most dramatic change in the behaviour of the pavement structure occurred when the moisture state and condition of the upper stabilized subbase changed.

It can be concluded that any remarkable change within the state of the upper weakly cemented subbase layer will be reflected by these indicators.

This is also true for resilient behaviour such as surface deflection (RSO) and radius of curvature (RC).

The indicators for permanent changes in behaviour are permanent deformation

(rut), shear strength (OCP) and dry density.

These indicators provide relatively valuable information, especially after a state of an excess porewater pressure existed within the upper subbase of pavement structure.

TRAFFIC CLASS (TRH4, 1985)

14

1000

800

600 rr' f

'f

RADIUS OF CURVATURE

400

200 o

1800

1400

1000

600

200

PWP

30

EPWP

10

20 30

I

I

EFFECTIVE ELASTIC MODULUS OF UPPER

STABILIZED SUBBASE

10

20

40 50

P

30

EPWP

DRY

40

PERMANENT DEFORMATION ON

ROAD SURFACE

50

--DRY

10 20 30 40

~

11..

u ....

cE

E

DCP UPPER STABILIZED SUBBASE

EPWP

I

DRY

40

10 20 30

I

DRY DENSITY UPPER STABILIZED SUBBASE

I

20

DRY:STAB-

DRY

EPWP

I-

40 III 10-E80.)5O

.' ORIGINAL

BASE AND STABILIZED

.

(VERY STFl'")

SUBBASES

.

CRACKED

.

(FLEXIBLE)

EPWP: EXCESS POREWATER PRESSURE

~

!

E P

SUBBASES CRACKED

..

IVE"Y FLEXIBLE)

50

FIGURE 3.-18

INDICATORS OF THE STATE OF THE PAVEMENT STRtlCTlJlftE

AT VARIOtlS STAGES OF TRAFFICKING AND MOISTtI!f£

COtJDITIONS AT MARIANNHILL

50

These indicators can be used to observe and quantify the state of the individual pavement layers at various stages of trafficking.

The pavement structure changed its state from an initially very stiff structure to flexible or very flexible depending on the loading and environmental conditions, such as accumulated water (EPWP), or high asphalt temperatures.

These paths of behaviour during the "life" of the structure are well defined, and are illustrated in

Figure 3.19.

The flow diagram in the figure is self explanatory.

Generally the pavement structure will change from very stiff to flexible and it is possible to change to very flexible if a state of excess porewater pressure does exist within the upper subbase layer.

If this condition is removed (drying of the layer) the structure can reverse to the flexible state, but with the upper subbase in a granulated (dry) state.

The asphalt layer will probably be fatigue cracked.

Provision should however be made to prevent the structure following any of the EPWP paths by adequate construction, drainage provision, regular inspections and timely maintenance (i.e. cracksealing, etc.).

The following section describes the state of the pavement defined including radius of curvature (RC).

The measured average road surface deflection (RSD) and radius of curvature (RC) for various stages of trafficking for measuring points 3 to 8 (dry test on Section 1) are given in Table 3.10.

The trafficking dualwheel load was

100 kN.

The HVS traffic was converted to E80s using a damage exponent· of d

=

4.

This value was used for convenience, and does not represent an accurate damaging value for this structure as was previously discussed (Section

3.2.3) •

PAVEMENT STATE

DEFLECTION

RUT

ASPHALT

WEAKLY

CEMENTED

SUBBASES

SELECTED

SUBGRADE

SUBGRADE

40AC

150C3

ASPHALT DEFORMATION;

CEMENTED SUBBASES

CRACKED INTO LARGE t

BLOCKS 1>41)

=

SUBBASE LAYER

THICKNESS

BAS 1- STRUCTURE No.

\-----.STIFF

ROAD CATEGORY

BITUMEN HOT-MIX BASE

I

FLEXIBLE

I

I

DRY

ASPHALT DEFORMATION;

CEMENTED SUBBASES

CRACKED INTO SMALLER

BLOCKS 1 31)

ASPHALT DEFORMATION

AND FATIGUE CRACKED.

CEMENTED SUBBASES

CRACKED IN TO SMALL

BLOCKS (21)

SURFACING AND BASE

CRACKED (FATIGUE)

CEMENTED SUBBASE

GRANULATED

BLOCKS 1<0,31)

FIGURE 3.19

DIFFERENT PAVEMENT STATES DURING THE HVS TESTS ON

THE PAVEMENT STRUCTURE AT MARIANHILL

The results given in the table are also illustrated in Figure 3.20.

The figure indicated an increase in RSD and decrease in RC.

After 5 x 106 E80s gradual, almost linear, changes in RSD and RC occurred.

As mentioned earlier the states of behaviour of pavement can be defined as very stiff, stiff flexible, very flexible in terms of surface deflection Freeme et aI, (1984).

The state of the pavement are defined in Table 3.11.

1,48

3,96

8,31

12,83

16,30

20,28

25,4

26,5

TABLE 3.10 - Average measured RSD and RC values at measuring points 3-8 on test Section 1 at various stages of trafficking

E80s

(x 10)6

Standard deviation (lJ.m)

RC (m) Standard deviation (m)

119

306

320

459

527

550

589

600

750

67

49

41

52

46

23

32

67

85

51(19)

255

227

250

150

987

484

540

276

289

217

296

194

58

99

61

36

93

178

137(86)

When these deflection boundaries are applied to the associated radius of curvature, it is possible to broaden the definition of the state of the pavement including radius of curvature, see Figure 3.21.

The appropriate radius of curvature and surface deflection for the different states are given in Table 3.12.

~ a: l.L.

a

~ a:

0

VI

::>

-1000

E

900

LLI a:

::>

I-

800

700

600

500

400

300

200

100

0

0,9

E

0,8

E

0,7 z

0,6

0 i=

(,)

0,5

LLI

..J

0,4 l.L.

0,3

LLI

Q

0,2

O,r-

0

0 1 234

I

STANDARD DEVIATION

5 6 7 8 9 10 II 12 13 14 15 16 1718 192021222324252627282930

NUMBER OF REPETITIONS (ME80s)

FIGURE 3.20

AVERAGE ROAD SURFACE DEFLECTION (RSDJ AND RADIUS

OF CURVATUREfRCJ UNDER A 40kN DUAL WHEEL LOAD

-1000 e

900

UJ

It:

;l

800

•.• 700

-

~ 800 -

It: a

500

~ 400

(J)

;l

300 o

200

~ 100 o

0,9 e

0,8 e

0,7

Z o

0,8

~ 0,5-

~ 0,4

..J

u..

0,3-

UJ o

0,2

0,1 -

00 I

STIFF (S)

VERY STIFF (VS )

XCESS PORE-

WATER

PRESSURE

FIGURE 3.21

AVERAGE ROAD SURFACE DEFLECTION (RSDJ AND RADIUS OF

CURVATURE (RCJ MEASURED UNDER A 40kN DUAL WHEEL LOAD

This section describes the development of a three dimensional model to describe the resilient behaviour of the structure, using the effective elastic modulus of the upper stabilized subbase, E, the road surface deflection under a 40 kN wheel load,

RSD, and the radius of curvature,

Re.

As indicated in

Table 3.2, Figure 3.15

and Figure 3.18, the road surface deflection is largely dependent on the effective elastic modulus of the upper weakly cemented subbase.

To construct the model it was necessary to obtain mathematical relationships between road surface deflection, modulus and radius of curvature.

The measured values used to obtain the relationships are given in Table 3.13.

TABLE of states of the pavement behaviour (After

Freeme et a1, (1984)

State Approximate deflection range (mm)

Very flexible

>0,7

Pavement behaviour predominantly controlled by very high IIIOdulus(>5 000 MPa). Cemented subbase layers acting as slabs.

Pavement behaviour controlled by ceaented subbase layers with high moduli (>3 000 MPa).

Subbase layers could be cracked but blocks tends to be larger than 1 m in diaaeter.

Pavement behaviour controlled by layers with reasonably high IIIOdul1(800 - 3 000 MPa).

Cementitious subbase layers cracked into smaller blocks.

Diameter of blocks 1,5 x layer thickness (t).

Pavement behaviour controlled by materials flexible in the granular state usually with low moduli «800 MPa). Subbase layers tends to be susceptible to excess porewater pressure.

TABLE 3.12 - Road surface deflection and radius of curvature in the different states of behaviour

State RSD

(nun)

RC (m)

Very stiff (VS)

Stiff (S)

<0,2

0,2 - 0,4

Flexible (F) 0,4 - 0,7

Very flexible

(VF)

>0,7

>700

350 - 700

150 - 350

<150

TABLE 3.13 - Measured and calculated values used to obtain the relationships

RSD

(llm)

*

RC (m)

**

E (MPa)**

180

350

500

600

746

1 670

1 000

450

270

250

150

50

7 428

2 200

2 200

800

550

260

* Results from Table 3.2

**

Results from Figure 3.10

In Figure 3.22, the relationships between RSD and E, and RSD and RC are given.

Confidence limits are also indicated on the figure, because the real mathematical functions are not purely hyperbolic.

The confidence limits were derived using the standard deviation of the constants, obtained from the products of RSD and E and RSD and RC, respectively.

These values are indicated in Table 3.14.

_ IBOO

11600 c en

0:: z

Q le.>

LtJ

~ lJ..

1400

1200

1000

LtJ c

BOO

LtJ e.>

<l: lJ..

600

0::

:::l en c

<l:

400

0

0::

200

\

,

'

,Ix

,

IJ \

\

\

\

\

\X

\

,X

,

"-

\

0

CD:

@:

@:

RSD. ""000

RSD= 731000/E

I}

RSD= 313000/E

RSD IN

(fL m)

E IN (MPa)

"

"-.leCD

~

® ........

--

------

3 4

5

-

-----..2-

6

-

7

2

EFFECTIVE ELASTIC MODULUS,E(MPa)

(II

10

3)

2000 -

LtJ c

LtJ e.>

<l: lJ..

0::

:::l en c

<l:

0

0::

E

3-

1800 c en

0::

1600 z

Q le.>

LtJ

1400

~

1200 lJ..

1000

800

600

400

200

~

I \

\

\ \

\

\

\ \

\ \

<0:

RSD. 17DOOO/Re}

RSD IN

(fLm)

@:

RSD= 136000/RC

RC IN

(m)

RSD= 1020001 RC

@:

\

\

\ \

X \

\ x,CD

"

•.•...

rJ),'X-

....•.

~

-

--

......-

--

-----

0

0

100 200

300 400 500 600 700 800 900 1000 1100 1200

RADIUS OF CURVATURE, RC

(m)

FIGURE 3.22

RELATIONSHIP BETWEEN ROAO SURFACE DEFLECTION,

ELASTIC MODULUS OF THE UPPER SUBBASE AND RADIUS

OF CURVATURE

RSDxRC (llm.m) RSDxE (llm.MPa)

(x103) (x103)

180,0

157,7

135,0

150,0

111,9

83,5

Average 136

Standard deviation +34

1 337,0

770,0

1 100,0

480,0

264,0

434,2

731

+418

The average values (constants in hyperbolic relationships)

136 000 and 731 000 were used to calculate both curves confidence limits, are calculated using the average values plus and minus the standard deviation, respectively.

From these calculations, the relationship between the RSD and RC appears to be more accurate than RSD versus E.

The accuracy of the relationships are indicated when the RSD is plotted against the constant divided by the independant variable,

Le.

E and RC.

See Figures 3.23 and 3.24.

The relationships between RSD and C/E and K/RC are indicated, where C and K are the different constants.

If the previous relationships were pure hyPerbolic, the latter relationships

RSD where C

=

7,428 x 10-1 C/E - 5,668

*

10-5 (C/E)2 •.•• (3.1)

=

1 149 000 (Upper Confidence limit)

=

731 000 (Average)

=

313 000 (Lower Confidence limit) with RSD in (llm)and E in (MPa).

2000

1900

1800

1700

1600

1500

1400

E

:l

1300

0

«

0 a:

0 en a:

Z g

Iu

W u..

w

0

1200

1100

1000

900 w u

«

800 u..

a:

700

:J en

600

500

400

300

200

100

RSD=7,428 IC/E)x 10-

1:-

5,668

X

10-sIC/E)2

WHERE C I/Lm. MPo)

CD:

1149000

®;

731000

®:

313000

0 500 1000 1500 2000 2500 3000 3500 4000 4500

FACTOR OF MODULUS,CI E I fl MPo)

FIGURE 3.23

RELATIONSHIP BETWEEN ROAD SURFACE DEFLECTION

(RSDJ AND EFFECTIVE ELASTIC MODULUS (EJ OF rHE

UPPER SUBBASE

1400 u..

w

0

w

u

.•..

1300 e

.5

1200

0 en

Q:

1100 z~

0

1= u

1000

900

800

700

Q:

::> en

600

0

<l

0

Q:

500

400

300

200

100

0

0

2000

1900

1800

1700

1600

1500

CD :

102000

®:

136000

®:

170000

500 1000 1500 2000 2500

FACTOR OF RADIUS OF CURVATURE,

K,/RC

3000

(1

1m)

3500

FIGURE 3.24

RELATIONSHIP BErWEEN ROAD SURFACE DEFLECTION

(RSDI AND RADIUS OF CURVArURE (RCI

and RSD where K

- 3.62

-

=

1,064 (K/RC) - 1,663

*

10-4 (K/RC)2 •..••••• (3.2)

=

102 000 (Upper Confidence limit)

=

136 000 (Average)

=

170 000 (Lower Confidence limit) with RSD in (~m) and RC in (m).

The regression analyses for above relationships are given in

Tables 3.15

and 3.16.

Dependent variable is RSD

Independent variable is C/E

1

2

.74280612

-.00005668

.19351785

.00007786

3.83843717

-.72791937

Intercept .•.••••••••.•••••

Multiple correlation ••.•.•

0.00000

:95839

Data No RSD

TABLE OF RESIDUALS

Observed RSD Estimated Residual

Std.

Resid.

4

5

6

1

2

3

180.000

350.000

500.000

600.000

746.000

1670.000

72.552

240.557

240.557

631.419

887.141

1640.424

107.448

109.443

259.443

-31. 419

-141.141

29.576

.555

.565

1. 339

-.162

-.728

.153

Dependent variable is RSD

Independent variable is'K!RC

1

2

1.06429973

-.00016630

.07931453

.00003170

13.41872250

-5.24536416

0.00000

.99451

.98905

71.01897

TABLE OF RESIDUALS

Data No RSD Observed

RSD Estimated Residual Std. Resid.

1

2

3

4

5

6

180.000

350.000

500.000

600.000

746.000

1670.000

141.669

306.465

493.898

529.764

828.256

1664.512

38.331

43.535

6.102

70.236

-82.256

5.488

.540

.613

.086

.989

-1.158

.077

The tables indicate relatively good multiple correlations, i.e. 95 and 99 per cent for RSD versus E and RSD versus RC respectively.

The standard errors of the RSD estimate are 193,7

~m and 71 ~m respectively.

The relationship between RSD and

K!RC appears to be fairly accurate.

In order to establish the three dimensional model, the relationship between E and RC was also studied.

This relationship appears to be a relatively accurate parabola.

Multiple correlation is 97 per cent with a standard error in

E of 586 MFa.

The relationship obtained is :

E

=

5,528 RC + 2,510

*

10-3 (RC)2 with E in (MFa) and RC in (m).

The regression analysis of E versus RC is given in Table

3.17.

The relationship is illustrated in Figure 3.25.

To develop the behaviour model only two of above equations are necessary, i.e. equation 3.2 and 3.3.

The RSD and E values are calculated from different RC values.

Equation 1 was left out on purpose because of the relatively high standard error involved.

In order to obtain a three dimensional plot of the three variables i.e. RSD, RC and E, a computer program, DISSPLA, available at CSIR computer centre was used.

The data input program and results are given in Appendix B.

In Figure 3.26

the computer plot of the model is illustrated.

As the original results include the EPWP state, the rapid change in resilient behaviour is well illustrated by the sharp curvature of the model in this state.

In the dry state the model indicates gradual change in resilient behaviour.

As previously discussed (Section 3.2.5), this pavement structure behaved almost linear elastic throughout, for normal loading.

Therefore it was decided to call the curved surfaces in the model the "ELASTIC SURFACE" and the plane parallel with the RSD, i.e.

orthogonal to the RC-E plane, the "ELASTIC WALL".

Any behaviour which does not coincide with the elastic line is therefore not linear elastic.

In other words, the behaviour on the elastic line indicates linear elastic behaviour.

In this pavement

-0

CL

::t

6500

6000

L&I j:

en

«

..J

en

5500

::l

..J

::l

0

5000

0

~ u

4500

4000

L&I

3500

L&I

~

I-

3000

•...

w

2500

2000

1500

1000

500

0

0

10000

9500

9000

8500

8000

1500

1000

E

=

5,528

RC+

2,5/0

X

16!(RC)2

E IN (MPo)

RC IN(m)

100 200 300 400

RADIUS

500

OF CURVATURE,

600

RC (m)

100 800 900 1000

FIGURE 3.25

RELATIONSHIP BETWEEN EFFECTIVE ELASTIC MODULUS

OF THE UPPER SUBBASE AND THE RADIUS OF CURVATURE

structure the dry "cut off" point appears to be at an RSD value of approximately 600 m.

(Dry cut-of is defined as the maximum deflection reached in the dry state on the road surface.

(See Figures 3.10 and 3.26.)

Dependent variable is E

Independent variable is RC

5.52787639

.00251021

1.09818279

.00142015

5.03365783

1.76756148

Intercept .••••••••..••••••

Mul tiple correlation •.••••

0.00000

.97361

TABLE OF RESIDUALS

Data

No

1

4

5

2

3

6

7

8

9

E Observed

8000.000

3500.000

3000.000

2500.000

2100.000

1900.000

1500.000

800.000

550.000

E Estimated

7901.381

3263.523

3717.030

1716.911

1807.211

1572.835

1384.176

1538.857

885.661

Residual

98.619

236.477

-717.030

783.089

292.789

327.165

115.824

-738.857

-335.661

Std.

Resid.

.168

.404

-1. 224

1.337

.500

.559

.198

-1.262

- .573

----------------------

-

.

The model, which is illustrated in Figure 3.26 describes the resilient behaviour during the normal loading (traffic) in the dry and excess porewater pressure states in terms of effective elastic modulus of the upper subbase, surface deflection and associated radius of curvature.

It can be used as an visual aid to understand the complex field of pavement behaviour.

The model can also be used to determine two of the unknowns if one of the three variables is known by entering the model with the known (measured or guessed) value on the appropriate axis.

The mathematical relationships can also be used for calculation purposes.

This part of Chapter 3 describes the main results obtained from HVS testing on a bitumen base pavement structure at

Mariannhill, in Natal.

The weakly cemented lime stabilized subbases were constructed of locally available weathered granitic material.

Permanent deformation and resilient behaviour were observed on three test sections, both in the normal moisture (dry) and excess porewater pressure states.

Moisture accelerated distress (MAD) was observed during the excess porewater pressure (EPWP) state within the upper subbase layer.

Good behaviour was observed in the dry state and it is expected that this type of structure will be structurally adequate to carry E4 traffic without major rehabilitation necessary.

It is however important to monitor pavement state in order to adequately design timely minor rehabilitation, such as cracksealing, drainage improvement etc.

According to mechanistic analyses, the horizontal strain values initiated at the bottom of the asphalt and lime treated layers appeared to increase non linearly, with a marked increase during the moisture accelerated distress

(MAD) phase.

During this phase fatigue distress of the asphalt and subsequent pumping from the upper subbase are possible.

Care should be taken in specifying cementitious subbase materials to be non-erodible.

(De Beer, 1985 (c»

The surface deflection and radius of curvature appear to be

relatively good indicators of behaviour of this type of structure, both in the dry and wet states.

Permeability tests on the recovered stabilized subbase layers indicated higher permeabilities at the bottom of the layer than at the top of the layer. This is believed to be a demonstration of the beam bending concept resulting in a higher degree of cracking at the bottom of these layers after trafficking.

DCP results indicated relatively high penetration rates within the upper subbase after trafficking, especially after the excess porewater pressure state.

It was also possible to quantify the state of cracking of the weakly cemented subbase layers.

The different states of behaviour of this pavement structure are described using a flow diagram which may be useful in future rehabilitation design.

It was also shown that the state of the pavement can also be defined including the radius of curvature.

A behavioural model of this pavement structure was developed using three parameters, i.e. road surface deflection, radius of curvature and the effective elastic modulus of the upper subbase.

It was indicated that mainly the effective modulus of the upper subbase governs the deflection on the road surface, especially during the excess porewater pressure state.

The results shows that the modulus, road surface deflection and radius of curvature are interrelated.

The model or the mathematical relationships can be used to obtain any of two unknowns in the model.

The objective of the HVS tests at this site was the same as that at Site 1 at Mariannhill.

However, the tests and analysis were hampered by a longitudinal settlement crack through all the layers.

(It finally appeared on the surface of the test section during HVS trafficking).

It is, however, be-

lieved that the results obtained will be very useful in evaluating the influence of such non-traffic-associated cracks on the future behaviour of this type of pavement design including weakly cemented subbase layers.

Slope instabilities often cause cracks in road structures in Natal (De Beer,

1983).

Two sections were tested at this HVS site.

Detailed evaluation of the test results at this HVS site is given elsewhere (De Beer, 1984(b».

The road structures as designed and as tested are illustrated in Figure 3.27.

The main differences between the two structures are:

(a) vertical crack through all the layers, including the fill material,

(b) soft and moist zone on the sides of the crack as well as a

+ 70 mm soft and moist sandy layer at the bottom of the lower weakly cemented (lime-treated) weathered granite subbase layer.

This layer was partially carbonated according to a chemical test (Netterberg,

1984).

At distances away from the vertical crack, the soft horizontal layer was also evident.

The vertical crack was discoloured and visible through all the layers in the structure.

The average permanent deformation of the road surface at various stages of trafficking on the two test sections is illustrated in Figures 3.28 and 3.29.

Figure 3.28 indicates a gradual increase in asphalt deformation on Section 1 in the dry state for the first 520 000 repetitions (20 ME80s).

AS TESTED

SETTL.EMENT

CRACK

ISOC3 BROWN WEATHERED

GRANITE

ISOC4 MOWN WEATHERED

GRANITE

SROWN WEATHERED i60 C2 GRANITE

100 C2 BROWN WEATHERED

GRANITE

70 G9 SOFT AND WET

F>ARTlAL.L.Y CARBONATED

L.AYER (SANDY)

• fZ3

WET UNCARBONATED

WET· F>ARTIAL.L.Y CARBONATED

• MATERIAL CODe: IN ACCORDANCE

FIGURE 3.27

PAVEMENT STRUCTURE AT FIGTREE HVS SITE

I , SATURATED

2' WET TO DRY z o

~ 25

2 a::

~ 20

"" o z

~ 15

+----+RUT

1_

DEPTH (STRAIGHT EDGE)

DOWNWARD CHANGE (PROFILOMETER)

ASPHALT DEFORMATION

ASPHALT TEMPERATURE (20_30

0

C)

!

u 10

......•

I

~--.J

:

~

I

I

.•...•.-+."

A·-·· ••

I r

I

~,.."..-..

.• __

._+

WATER INGRESS: SURFACE

.

(THROUGH LONGITUDINAL

CRACK ONLY)

DEFORMATION FROM SUBBASE LAYERS

ASPHALT TEMPERATURE (20.

30°C

I

400000 600000

NUMBER OF REPETITIONS

FIGURE 3.28

AVERAGE DEFORMATION OF THE ROAD SURFACE AT VARIOUS

STAGES OF TRAFFICKING ON SECTION' ATINDICATED WHEEL LOAD

+----+

RUT DEPTH (STRAIGHT EDGE)

_.

DOWNWARD CHANGE (PROFILDMETER)

40 kN (DUAL-WHEEL

LOAD)

150 kN(SINGLE-WHEEL

LOAD)

400000 600000

NUMBER OF REPETITIONS

FIGURE 3.29

AVERAGE OEFORMA TION OF THE ROAD SURFACE AT VARIOUS

STAGES OF TRAFFICKING ON SECTION

2

AT INDICATED WHEEL LOAD

Traffic loading

(kN)

Actual repeti-

Dry

Dry

20-30

20-30

520 000 Total section tested (8 m)

3 108 Subsurface water ingress, both sides of the test section,

(one half of the section)

3 108 Water ingress mainly from the surface through the longitudinal crack in the asphalt

156 382 Subsurface water ingress every alternate 24 hours both sides of the test section (4 m)

Total number of

679 490 repetitions applied to test section

280 494

17 255

Total number of

297 749 repetitions applied to test section

"Dry" is indicative of the normal in-situ moisture condition in the sublayers.

« optimum moisture content)

"Saturated" in- this thesis is indicative of a condition where excess porewater pressure exists within the structure of the test section, especially between the asphalt and upper subbase layers.

Singlewheel load 150 kN, tyre pressure 1448 kPa and dualwheel loads 40, 100 kN, tyre pressure 690 kPa.

After 520 000 repetitions the average rut was 9 mm.

After the introduction of water the rut increased to 20 mm and

*

A decrease in the rate of deformation was observed after the dualwhee1 load was decreased from 100 kN to 40 kN, and after the application of water was varied.

Figure 3.29 indicates a relatively low increase in asphalt deformation after 280 000 repetitions (E80s) on Section 2.

The deformation increased to an average of 2 mm in the dry state with 40 kN dual-wheel load trafficking.

The dualwheel load was changed to a singlewheel load of 150 kN and the deformation increased to an average of 25 mm after a further

17 255 repetitions.

After studying the origin of the rut, it was concluded that more than 60 per cent of the rut originated in the asphalt layers.

Similar behaviour was found at the previous HVS site (Mariannhill) where more than

80 per cent of the rut originated from the asphalt layers.

(Testing temperature between 20 DC and 30 DC).

The average RSD and RC under a 40 kN dualwheel load on

Sections 1 and Section 2 are illustrated in Figure 3.30.

It is important to note that both saturated conditions and excessive loading, even in the dry state, resulted in an increase in the deflection and a decrease in the RC.

With normal trafficking, however, it is believed that this structure will not reach the very flexible state under dry conditions.

Pressurized water (1 to 2 m) was introduced into the sublayers of the structureon both sides of the test section, for half the length of the test section.

a

"-

0 en

:;)

0

<[

II::

~

~

~IOOO

0

0,6

~ z

0,4

0 i= u

!oJ

..J

"-

!oJ

0

0

---

----'

~ 150 kN

I

I

I

••

,

I

SECTION'

- - SECTION 2

400000 600000

NUMBER OF REPETITIONS

FIGURE 3.30

AVERAGE RSD DEFLECTION AND RADIUS OF CURVATURE (RCI

UNDER A 40kN DUAL- WHEEL LOAD ON SECTIONS 182

It is, however, important to seal any cracks on the surface, including the settlement cracks found on this route, to prevent water ingress and rapid carbonation of the weakly cemented layers.

The average MOD depth deflections at various stages of trafficking on Sections 1 and 2 are illustrated in Figure 3.31.

The figure indicates that at the end of the dry state more than 60 per cent of the surface deflection originated in the selected layer downwards on both test sections.

Less than

10 per cent originated in the asphalt layers.

The remaining

30 per cent of the deflections originated in the weakly cemented subbase layers.

However, the deflections were small.

In the saturated state the contributions to the surface deflection changed from 60 to 50 per cent in the selected layers downwards, 10 to 15 per cent in the asphalt layers, and 30 to 35 per cent in the stabilized subbase layers.

These results were similar than those found at

Mariannhill.

From the HVS results of tests on Section 1 an average permanent deformation of 20 mm in the asphalt is not expected within the design life of this pavement; however, 10 rom is expected after approximately 20 ME80s

(!

25 years after the opening of the road).

It is important to note that after 10 years the weakly cemented subbase layers will be cracked and possibly fractured.

The HVS, however, proved that fatigue cracking will not occur in the asphalt layer during the design life of this road structure if the subbases remain dry.

Early fatigue cracking will be experienced in both the weakly cemented subbase layers, especially the upper layer.

However, this will not result in immediate functional failure of the

E

E

500

Q

C

0

C

C

;:)

III

III

U

C

...

:it

1000

III

Q

..

%

III

III tI.

1500

0

0 0,1

DEFLECTION lmm'

1,1 1,2

0

0 o

ME80.( INITIAL!

4 MEBO.

12,5 MEBO.

16,4 ME80.

20,3 MESO.

E

E

Q

C

0

C

~

0

~

III

III

III u

C

•••

C

;:)

III

% to

L

III

0 llOO

1000

DEFLECTION (mml

20,6 ME80.

(SEMI-SATURATElll

20,4 MESO.

(SATURATED' lit

C

;:)

U

C

...

III

E

E

llOO

~

3

III

III

Q

C

0 a:

1000

%

~

~

IllOO

0

DEFLECTION (mmI

0,6

0,9

10 E80.

107482

E80.

21111920 E80.

280494

E80.

1,0

W

-..J

())

I

FIGURE 3.31

AVERAGE MDD DEPTH DEFLECTIONS AT VARIOUS STAGES OF TRAFFICKING ON SECTIONS' 82

structure (relatively short precracked phase, mainly because of the longitudinal crack).

From the limited HVS test results it appears that the weakly cemented (uncarbonated) weathered granite material used in this structure behaves relatively better than the weathered granite used at the previous site (Mariannhill), although the effective elastic moduli values for the stabilized subbases were relatively low.

The effective elastic moduli values of the upper subbase varied between 10 000 MPa and

70 MFa and for the lower subbase between 114 MFa and 30 MFa, respectively.

It is believed that the reason for the very low moduli values is the existance of the settlement crack in the cemented layers (De Beer, 1984 (b))• The weakly cemented granite at Mariannhill appeared to be more sensitive to water (EPWP) than the granite material at this site.

It is important to note that the road structure at Mariannhill consisted of well stabilized subbase layers and excellent construction without thick and soft interlayers, carbonation or settlement cracks.

There were virtually no differences in ultimate behaviour during the dry states of testing on these two sites.

It is believed that the weakly cemented weathered granite material at this site (Figtree), built according to the design (no interlayers, carbonation and/or settlement cracking), will be able to carry the design traffic over the design period (20 to 30 ME80s over

20 years).

Lime or cement may be injected into the settlement cracks to retard the carbonation and to decrease the water sensitivity of the subbase material in the vicinity of such cracks.

The cracks must also be sealed to prevent water and air from penetrating into the road structure.

More care should be taken in corstructing cemented layers to avoid uncemented soft and wet interlayers, since these cause a dramatic reduction in the fatigue life of the stabilized and asphalt layers of this design.

Quality control should include the carbonation test proposed by Netterberg, 1984.

On fill foundations cementitious and asphalt layers should be built only after most of the primary consolidation and

settlement have taken place to avoid detrimental cracking in the main structural layers of these road structures, including weakly cemented subbase layers.

It can be concluded that the existance of the vertical crack and the soft zone influenced the measurement of the relative deflections and hence the calculation of the effective elastic modulus of the weakly cemented subbase layers.

This leads to very short fatigue lives of the asphalt layers.

The origin of the rutting occurred within the soft zone in the vicinity of the crack and not from the weakly cemented subbase layers, (see Figure 3.32), therefore this type of weakness within these structures must be minimized.

At this site the two weakly cemented subbases consisted of lime treated Berea Red sand material.

The objective of these tests was also to study the behaviour of the subbase layers again with and without the introduction of water into the structure.

Detailed evaluation of the test results at this

HVS test site is also given elsewhere (De Beer, 1985(a».

The design of the road structure was the same as that of the structure at Mariannhil1 (Figure 3.1), except that the aspha1t base layer was 90 mm thick.

The structure was tested without the asphalt surfacing layer.

The weakly cemented subbases were constructed of fine grained weakly cemented (lime treated) Berea Red Sand material.

Two sections were tested at this HVS site, each in the dry and saturated states.

TRANSVERSE

0,0 0,2 0,4 0,6 0,8

Iii f ii'

DISTANCE

(m)

1,0 1,2 i i

1,4 1,6 1,8 i

i i

LOWER SUBBASE

(L.TSB

(2»)

SOFT AND WET

PARTIAL.L.Y

CARBONATED L.AYER

BEREA REO SAND

SEL.ECTED

L.AYER

L.IME

TREATED

WEATHERED

GRANITE

FIGURE

3.32

FINAL rRANSVeRSe PROFILe

OF' seer/ON I WHeRe SURFACe

AND SUBSURFACe wArER WeRe INrRODUCeD

The traffic loading and moisture conditions differ for each test and are summarized in Table 3.19.

No trafficking was done on the control section, but deflection measurements were taken to supplement the main tests, although measurements were more limited.

1

2

Traffic loading

(JcN)

Asphalt teJDperature

(Oe)

Actual repetitions

Moisture* conditions

80

100

100

40

40

40

17 - 25

20 - 27

20 - 27

20 - 27

17 - 30

17 - 30

201 617

3'l8 383

2 067

45 419

263 000

106 674

Dry

Dry

Excess porewater pressure

Excess porewater pressure

Dry

Excess porewater pressure

*The moisture conditions refers mainly to the moisture condition of the two weakly cemented subbases, especially the interface between the aspba1 t and upper subbase.

Because the test sections are situated on a relatively steep slope, the total length of the subsurface of both sections was influenced by the excess porewater pressure (EPWP) during HVS trafficking.

Figures 3.33 (a) and (b) give the average deformation (rut) of the road surface at various stages of trafficking

Sections 1 and 2, respectively.

Pressurized water on

(1 to

2 m) was introduced after approximately 15,6 MESOs on

Section 1 and 263 000 ESOson Section 2.

Hardly any rut development occurred in the dry state on either test section.

In the state of excess porewater pressure, especially in the interface between the asphalt base and the top of the upper weakly cemented subbase, moisture-accelerated distress (MAD) The rate of change in deformation appeared to be linear and varied between 300 and 500 mmper MESO. Typical HVSlongitudinal fatigue cracking appeared on the surface of the asphalt layer on the sides of both sections during the MAD

Fine subbase material was pumped from the weakly cemented subbase through these cracks and at some MOD Hardly any other fatigue cracks were evident on the surface of the asphalt base layer of either section after the tests.

Figures 3.34(a) and (b) give the RSD responses for the tested sections together with the associated RC.

On both test sections a gradual increase in deflection occurred in the dry state.

pressure state,

At there the start were of the mean surface excess porewater deflections of

0,35 mmand RCof approximately 400 m on both test sections.

Neither figure indicates any drastic change in the abovementioned parameters during the excess porewater pressure state.

On Section 1 the deflection increased to 0,40 mm only.

Both sections appeared to be in the stiff state after the introduction of excess of porewater pressure.

The relatively poor correlation between the surface deflection measurements and subsequent HVSresults shows that deflection and RC cannot, in this case, be used as a strong indicator of performance during excess porewater pressure states in this structure.

DRY

-

EXCESS POR£WATER PRESSURE z

-

E 55

E

~ 45 i=

<l

~ 35 a::

~

w

25 o

-

SOkN 100 kN

w

Cl

:i

5

:r: 0 u -5

,

.

1

.

.

o

3 6 '3 12 15 IS 21 24 27 30

NUMBER OF REPETITIONS (MESOsl

«a

J

SECTION t

E

55

.§ z

45 o i=

~ 35a:: -

~ 25o

;!';

15-

w

~

<l

:r: u

5o

DRY

100000 200000 300000 400000

NUMBER OF REPETITIONS (EBOsl

500000

« b

J

SECTION 2

FIGURE 3.33

AVERAGE DEFORMATION (RUT) OF THE ROAD

SURFACE AT VARIOUS STAGES OF TRAFFICKING

ON THE TWO SECTIONS

!

2000

III

~ 1600 c

~ 1200

::>

...

o

~ 400 o

<II

It:

0,4

E

!

z

0

;::

U

..J

I.

III

0

0,3

0,2 -

0,1

0

0

3

2-

6 9 12 15

NUMBER OF REPETITIONS

IS 21

(MESOsI

(a

J

SECTION f

24 27 30

TRAFFICt<1NG WHEEL LOAD! 40 kN

EXCESS PORE -

I

WATER PRESSURE

100000 200000 300000

NUMBER OF REPETITIONS lE80'1

400000

FIGURE 3.34

AVERAGE RSD AND RC UNDER A 40kN DUAL-

WHEEL LOAD ON 'HE ,WO SEC II ONS

Figures 3.35(a) and (b) give the average MOD depth deflection under a 40 kN dualwheel load at various stages of trafficking on the two· sections.

The figure indicates a well balanced pavement structure in terms of depth deflection, even in the MAD state.

The stiffness of every layer tends to decrease with trafficking (MOD depths represent layer interfaces).

The major part of the surface deflection is attributable to the selected layer downwards (405 mm and below), which contributes some 75 per cent of the total surface deflection.

In the case of Mariannhill and Figtree the percentages were approximately 50 to 80 and 65, respectively.

Visual changes on both test sections were almost the same in the dry and excessive porewater pressure states.

In the dry state hardly any change was noted on either of the sections.

A rapid change in permanent deformation with minor surface fatigue cracking was evident on both sections in the excess porewater pressure state.

Figure 3.36

shows a profile of Section 2, taken after completion of testing, at the position of maximum rut.

The major part of the permanent deformation inside the test section was due to the loss of weakly cemented upper subbase layer material.

The material was pumped towards the sides of the test section between the bottom of the asphalt and the top of the weakly cemented subbase layer.

The failure mechanism was clearly indicated by the horizontally displaced material (pumped layer).

The same mechanism of failure was evident in another seven similar profiles inspected on the two sections after HVS testing.

The displacement of the top of the upper layer was attributable to this layer's lack of resistance to erosion in the excess

0

0

E

E

0

«

0

0: l&J

...

l&J

0

:r

...

ll.

~

9 l&J lD l&J l)

«

0:

:::I en

500 rooo

1500

0,2

DEFLECTION lmml

0,4

0,6 0,9

ME9O.

0

I,'

3,2

1t,3

'41,441 ll.

l&J

0

E

.5

l&J l)

«

IL.

0:

:::I en

0

«

0

0: l&J

:r

...

~

0

J l&J lD

1000

1500

500

0

0

DEFLECTION lmml

,

0,2 0,4 0,6

I

x

D o

0

I

I

I

rl

I I

X D

a

E80s

10

2014117

JI

3574241

3.,4174

0,8 1,0 1,2

2000

( a I SECTION t

2000

«b

J

SECTION 2

FIGURE 3.35

AVERAGE MOO DEPTH DEFLECTION UNDER A 40kN DUAL-WHEEL LOAD

AT VARIOUS STAGES OF TRAFFICKING ON THE TWO SECTIONS

.

W

00

-..J

o

0,1 0,2 0,3 0,4 0,5 0" 0,7 0,8 0,9 ~O 1,1

DEoPTHI:nmi i

F-T-~-S-;-S-~-C-;-IO-~-:t_-~~~~:o----l

I,!! 1,6 1,7 1,8

-,

_1---':

, i r'II

PUMPED LAYER (DUE TO

CARBONATED LAYER

UNSTABILIZED LAYER

SELECTED SUBGRADE

LAYER

FIGURE 3.36

PROFILE OF HVS TEST SECTION

2

TAKEN AT POINT /I

AT UMGABABA

porewater pressure state.

It is worthwhile to note that the pumped material as well as the top of the subbase were uncarbonated before and during the testing.

It is proposed that a durability test should be developed to quantify the erodibility of cementitious material.

Work in this regard has already been started at the NITRR. (De Beer,

1985 (c»

Because of the different failure mechanisms encountered on this HVS site, the limitations in the current mechanistic design method are realised because durability criteria for weakly cemented materials are not considered at this stage.

It was decided to include in this thesis the test results of this HVS site because very important behavioural characteristics were found.

In this case the design was the same as for the previous three sites, i.e. asphalt base of approximately

100 mm thick, asphalt surfacing 40 mm thick on top of two weakly cemented subbase layers, of 150 mm thick each.

The subbases consists of blended river gravel and sandstone, treated with 1,5 per cent lime and 1,5 per cent slag.

Because this was a rehabilitation project the sandstone of the old road was re-used together with new river gravel material.

The old asphalt was also re-used (recycled) with a ratio of 70 per cent virgin to 30 per cent new conventionally continuously graded asphalt.

The recycled (hot processed) asphalt was layed separately in two layers of 65 mm and 45 mm, respectively.

Detailed analyses and evaluation of this HVS test site is given elsewhere (De Beer, 1985(b».

Inspection test pits prior to HVS testing revealed that the weakly cemented upper subbase was not in a cemented state but rather in a granular state.

In another test pit the upper 50 to 75 mm of the upper subbase was relatively strongly cemented.

In both test pits, however, the lower subbase was

also relatively strongly cemented and probably of C3/C2 quality.

The main objective of the HVS testing at this site was to evaluate the structural behaviour of the recycled asphalt base and secondly to predict the future behaviour of the rehabilitated pavement.

Because of the type of design (i.e. weakly cemented subbases) the behavioural characteristics in view of this thesis is of utmost importance because relatively early fatigue distress and excessive rutting of the asphalt layers occurred after HVS trafficking.

This distress was shown to be directly a function of the support from the upper subbase both in the dry and in the excess porewater pressure (EPWP) state.

In the dry state relatively early fatigue distress occurred.

When water was introduced (EPWP state) after the cracking occurred, excessive rapid change in permanent deformation

(rutting) and pumping occurred.

The construction of the rehabilitated pavement was completed during 1981/82 and HVS testing was done from January 1983 to

July 1983.

The overall evaluation included Lacroix deflection and Dynamic

Cone Penetrometer (DCP) surveys in the slow lanes (both directions) of this route.

The three HVS test sections were selected on the basis of these results.

One section was representative of the weaker 10 to 15 per cent of the structure.

*

The other two sections represented the remaining

85 to 90 per cent.

The total length of the rehabilitation project was approximately 24 km of which 17,5 km included recycled asphalt in the base.

The other 6,5 km included conventional asphalt in the base layer.

The substructure was virtually the same for both parts of the route.

Based on the length of the route with relatively high deflections and high DCP values.

Section 1 was representative of the weaker 10 to 15 per cent of the route.

The pavement structure of this section is shown in Figure

3.37(a).

A test pit dug near Section 1 before the HVS testing and trenches excavated across the test section after the HVS test confirmed that the upper subbase was not weakly cemented and rather in a granular (poor) state.

The granular state was first suspected after DCP measurements of over 5 mm per blow in this layer.

The lower subbase was relativele stronger cemented and remained in a solid state after the HVS testing, without major cracking or granulation.

The section was trafficked with a 70 kN dualwheel load and fatigue cracks appeared on the surface after about 58 800 repetitions.

After 60 000 repetitions a well defined crocodile crack pattern was established.

After 280 000 repetitions the surface was extensively cracked, and at this stage the temperature of the asphalt layer was increased, by means of heaters, from an average ambient level of 25°C to

45 - 50 °C.

At this temperature permanent deformation occurred mainly the semi-gapgraded asphalt surfacing.

The rate of increase in deformation, however, was not critical when compared with that in other similar tests done in Natal (Freeme et al,

1982).

The measured permanent deformation versus the number of 70 kN load repetitions is shown in Figure 3.37(b)

·6·:·:·'?·

EG4 :;.;

:.Q';.

t50mm WEAKLY CEMENTED U••••ER SUBBASE

,~\~O·~

IN A POOR STATE IGRA:NULARI

( 0)

PAVEMENT STRUCTURE: WEAKE R 10 - 15% OF ROUTE

(HVS TEST I)

50

45 I-

40 I-

35 l-

E

E

30 l-

Z 251o

~ 20 I-

::E

~ 151ll.

~ 10 r

SURFACE

CRACKS

IFATIGUE I

~SPHALT

ASPHALT

TEMPERATURE

20-30·C

TEMPERATURE

40-50·C

100 150

RE"ETITIONS,

200 250

70 kN 1.,0

1,

300

FIGURE 3.37

PAVEMENT STRUCTURE REPRESENTATIVE OF

THE WEAKER

10-15"

OF THE ROUTE TO-

GETHER WITH THE PERMANENT DEFORMATION

AND CRACKING BEHAVIOUR RESULTING FROM

THE HVS TEST UNDER DRY CONDITIONS

The HVS test results from Section 1 and a mechanistic evaluation indicated that.fatigue cracks would appear on the weaker 10 to 15 per cent of the route within the following two to three years.

No excessive permanent deformation is expected if the structure remains dry.

The elevated temperature test showed that high temperatures in the asphalt under heavy traffic loading would not be critical with regard to permanent deformation of the asphalt surface and base layers.

Sections 2 and and 3 were representative of 85 to 90 per cent of the route.

The fatigue cracking behaviour of the asphalt layer of

Sections 2 and 3 was studied further at normal ambient temperatures.

Water was also introduced to study the reaction of the pavement structure to excess porewater pressure (EPWP) in the subbase of Section 2.

A test pit dug between Sections 2 and 3 before the HVS testing and trenches excavated across Section 2 after the

HVS test showed that the upper subbase was in a better condition than that on Section 1.

The pavement structure at

Sections 2 and 3 is shown in Figure 3.38(a).

The upper SO to 75 nun of the upper subbase was strongly cemented

(e2).

The rest of the layer was weakly cemented and in a granulated (poor) state, as was found on Section 1.

On Section 2 a 70 kN dualwheel load was used for the first

100 000 repetitions, after which the load was increased to

100 kN for the remainder of the test.

Fatigue cracks appeared on the surface after a total of 301 000 repetitions.

(See Figure 3.38(b».

40 mm W£A"ING COU"S£: $f:MHlAP-GRADED

"ECYCLIED ASPHALT

CONTINUOUSLY G"ADED

BASIE LAYER

I a) PAVEMENT STRUCTURE: REMAINING 8S-90 -;.

OF

ROUTE I HVS TEST 2 AND 3)

-

II

A

-

-

TEST rCTION

B

I I

SURI'ACIE AND SUBSURI'ACIE WAn"

SURI'ACf: WATE"

C'

I ON A

100 kN

MOISTUR e-ACCELf: RATErl

'·~~l

14'1

ON B

Y

t

II.,

P

A

~URI'ACf:

CRACKS

Cl'ATIO~~

FIGURE 3.38

PAVEMENr SrlWCrURE REPRESENrArlVE Or ~.gO"

Or rHE RourE rOGErHER WlrH rHE PERMANENr

DErORMArlON AND CRACKING BEHAVIOUR RESULrlNG

F'ROMrHE H VS rEST'. UNDER DRY AND WEr CONDlrlONS

Section 3 was first trafficked with a 40 kN dualwheel load for 800 000 repetitions and then with a 100 kN load for the remainder of the test.

After 200 000 repetitions of the

100 kN load, fatigue cracks were visible on the surface.

The fatigue response of the pavement structure at Sections 2 and 3 was better than that found at Section 1.

The HVS tests showed that fatigue cracks should not be experienced on the rest of the route (which is equivalent to Sections 2 and 3) within the next ten years or so with normal traffic

(E80 growth rate of 10 per cent).

Mechanistic analysis was used to predict a fatigue-crackfree period of 10 to 15 years.

(See Section 3.5.5 later).

The influence of the ingress of water into the subbase layers, after the occurrence of fatigue cracks in the asphalt base and surfacing layers, was also studied.

Water was sprayed on the cracked Section 2 and then subsurface water was introduced into the subbase layers.

The introduction of water into the pavement structure initiated moisture-accelerated distress (MAD) (see Figure 3.38 (b» , which demonstrates the necessity of preventing the ingress of water into this pavement.

The excessive permanent deformation that followed was mainly due to loss of fine material, through pumping (EPWP), from the upper subbase layer.

The permanent deformation and cracking on Section 3, were similar to those experienced on Section 2 in the dry state.

No water was introduced into this section.

With the ingress of water under normal traffic loading the results on Section 2 showed that the terminal level of permanent deformation (20 rom) would probably occur within four years, and within two to three years if the subdrainage was poor.

Sealing of the asphalt surface with a proven waterproof seal as soon as the cracking appears was strongly advised.

Local maintenance to improve the subdrainage was also advised in areas with poor subdrainage.

Section 2 was initially trafficked with a 70 kN, dualwheel load as was done on Section 1.

Because of the relatively better (strongly cemented) upper 50 to 75 mm of the upper subbase, no crack pattern was established after 100 000 repetitions in the dry state.

It was then decided to increase the wheel load to 100 kN.

A similar fatigue crack pattern as on Section 1 was established after

216 000 repetitions with the 100 kN wheel load.

The permanent deformation at that stage was approximately

3 mm.

See previous Figure 3.38.

During this stage of trafficking the fatigue crack length was measured regularly to establish a crack growth curve for this pavement structure.

The crack length (CL) is defined as the total length of cracks on the surface of the test section, in metres.

The crack growth is illustrated in Figure

3.39.

The figure illustrated the crack length (CL) versus number of actual repetitions.

After approximately 316 000 total repetitions, 50 m of cracks were measured.

A marked increase in crack length (125 m) occurred in a very short period of trafficking (~3 000 repetitions), after which the rate of increase in crack length decreased.

A final length of approximately 200 m of crack was measured on the test section.

It can be concluded from this information that rapid fatigue crack growth from the bottom of the recycled asphalt base layer to the surface will occur.

As previously mentioned water was continuously sprayed on the surface of the cracked section using a spray bar, in order to simulate normal rain conditions on this pavement.

Virtually no rapid rut development occurred for at least

100 000 repetitions (100 kN) in this state, alghouth slight pumping was observed.

After a total of 416 000 actual repetitions, pressurized subsurface water was introduced, using a set of inlet pipes both side of the

UI l&I

250 l&I

::E 200

!': oJ u

150

C) z l&I oJ

~

U cl[

II:

U oJ

0:(

•..

'00

50

X

,/

DRY •• t

X

I

X

I~

~X-

~URFACE AND

SUBSURFACE

WATER (EPWPI

_x

FIGURE 3.39

FATIGUE CRACK GROWTH ON THE ASPHALT SURFACE OF TEST SECTlON2

AT VARIOUS STAGES OF TRAFFICKING WITH INDICATED DUAL WHEEL LOADS

test section.

This was done only at one half of the length of the test section (See part B, Figure 3.38).

Excessive pumping started at 416 000 actual repetitions over the total section with increasing effect on where subsurface water was also introduced.

The test was ended with an average rut depth of 45 mm on part B and 15 mm on part A.

(See previous Figure 3.38)

The final crack pattern observed on the section is illustrated in Figure 3.40.

Part A experienced a crack pattern consisting of relatively large blocks.

The average diameter of the blocks was approximately 150 mm which was also the total thickness of the asphalt layers.

This block size was called the A-blocks.

On this part of the section only surface water was applied during trafficking.

The average rut depth at the end of the test was approximately 15 mm on this part of the section.

The maximum rut depth however was 65 mm at point 6.

At point

2 a local rut of 40 mm was measured.

See Figure 3.40.

On part B of the section, two block sizes were mainly observed.

On two local areas, points 8 to 11 and points 13 to 16 on the left of the section (facing the section from point zero), the A-block size (average diameter approximately 150 mm) was observed.

The rest of the section experienced smaller blocks with average diameter of approximately 90 mm.

This was called the B-blocks.

Maximum rut depths were established on two of the A-block areas, viz: 90 mm and 70 mm at points 9 and 14, respectively.

Where the smaller blocks were formed the average rut depth was less than 40 mm.

A local maximum rut of

55 mm, however, was measured at point 9.

Inspection test pits after testing revealed that the fatigue cracks were through to the bottom of the asphalt base layer in the areas consisting of A-block sizes.

In the area (part B) receiving subsurface water excessive

PART A

UPPER HALF: SURFACE WATER LOWER HALF: SURFACE

WATER

+

SUBSURFACE

X

SUBSURFACE WATER

+

MOD

• MAX.

RUT

==::A-BLOCKS:

--

AVERAGE DIMENSION 150 mm

(UPPER SUBBASE FAILURE I

B-BLOCI<S : AVERAGE DIMENSION 90 mm (WEAl< RECYCLED BASE LAYER fAILURE I

FIGURE 3.40

FINAL CRACK ANO RUT PATTERN

OBSERVEO ON SECTION 2 AFTER

WATER

APPLICATION ANO HVS TRAFFICKING

pumping occurred vertically through these cracks from the granulated upper subbase layer, hence the large permanent deformations of 70 mm to 90 mm.

In the areas consisting of the B-blocks, the fatigue cracks were not through the lower 50 to 75 mm of the recycled asphalt base layer, but extended only approximately 90 mm from the surface to within the recycled base layer.

As a result virtually no pumping from the upper subbase occurred in this area.

It is further interesting to note that the upper subbase at this position was in a relatively strongly cemented state.

It is believed that the influence of the stronger upper subbase support in this case was such that weaknesses within the recycled asphalt base layer were accentuated, hence the smaller blocks and fracturing of the asphalt.

This distress occurred a depth approximately equal to the average diameter of the blocks (90 mm) on the surface.

It is further interesting to note that it appears that the average diameter of the blocks on the surface, owing mainly to fatigue, approaches the depth of the weaker layer.

In this case two "weak layers" were identified, viz: the upper subbase (granulated) under the A-blocks, and secondly a weak sone wi thin the asphalt base layer where the B-blocks were observed.

In each case the "weak layer" was situated at a depth equal to the average block size on the surface.

If the block diameter approaches the layer thickness, "rocking" of the block occurs, without further fatigue distress because the blocks are simply too small to bend.

This block size was called the "optimum block size", which is actually the smallest diameter which could occur owing to fatigue distress.

A cross section of

Section 2, illustrating above mentioned failure mechanisms, is illustrated in Figure 3.41.

The figure is self explanatory.

Because the relatively poor state (granular) of the upper subbase at this site was known prior to HVS testing, it was decided to make a test pit through all the layers in the structure, approximately 800 mm away from the one end

HVS MOVING LOAD

+

~ f

145mm

~EMElnED

UPPER PART OF

UPPER SUBBASE

SURFACE WATER

-SURFACE WATER DID NOT ENTER THE WEAK

ZONE IN THE SAME MANNER AS THE OTHER

HALF OF THE SECTION.

LESS WATER ENTERED THIS LAYER,

LESS PORE PRESSURES DEVELOPED, LESS

FAILURE TOOK PLACE IN THE WEAK ZONE.

THUS BIGGER BLOCKS FORMED, PUMPING

FINES FROM THE WEAKER SUBBASE.

- WATER ENTERED WEAK ZONE FROM

TOP AND BOTTOM.

RESULTANT PORE PRESSURES AND LOAD

WEAKENED THE WEAK ZONE, FORMING THE

SMALLER BLOCKS OWING TO FATIGUE

DISTRESS WHERE THE UPPER SUBBASE

HAD THE HIGHER BEARING CAPACITY.

FIGURE 3.41

THE TWO FAILURE MECHANISMS EXPERIENCED

SECTION

2

ON HVS TEST

40mm

WE ARING COURSE

IOOmm

RECYCLED BASE

(2 x 50mm)

WEAK ZONE:

(TOTAL FAILURE

OF THIS 50mm

ASPHALT LAYER)

of Section 2.

See Figure 3.42.

This was done in order to study horisontal pumping resul ting during the excess porewater pressure (EPWP)state in Section 2.

A substantial amount of fine upper subbase material was horizontally pumped and was observed in this test pit during the test.

Fine subbase material was flowing out of the weak zone within the recycled asphalt base layer as well as from the upper subbase.

This observation is indicative of the enormous pressures which develop during traffic loading in the road structure when free water is allowed to be within relatively poor layers.

The road surface deflection (RSD) and radius of curvature

(RC) at various stages of trafficking measured on Section 1 are illustrated in Figure 3.43(a).

The average initial RSD was 0, 4S nun and increased to approximately 0,61 nun at the time of the appearance of fatigue distress on the surface of the asphalt (after approximately 60 000 repetitions, 70 kN dualwheel load).

The final RSDwas approximately O,SO nun after 370 000 actual 70 kN repetitions, which is 3,5 x 106

ESOs if a damage exponent d=4 is used.

A relatively low initial RC of 110 m was calculated from the RSDmeasurements.

The final RC was approximately SO m.

These relatively low RC values (and high RSD) are indicative of the relatively weak initial state of this pavement structure at

Section 1.

According to the previous Tables 3.11 and 3.12

this structure varies from flexible (F) to marginal very flexible (VF) in the dry state.

The effect of high temperature in the asphal t layers with respect to RSDand RC is negligible.

TEST PIT:

BOO mm FROM SECTION

END AT POINT 16

SUBSURFACE WATER

INTRODUCTION

7

BOOmm

I

II llllllll II

I

PUMPED FI.~ES

ON TH I S END FACE o

111l1l1l1l1l1l16

SECTION 2 TEST

PIT

HORIZONTAL CRACK

IN THE RECYCLED

BASE UPPER LAYER

(WEAK ZONE'

HORI ZONTALL Y

PUMPED FINES

ORIGINATED FROM

THE UPPER

SUBBASE

40mm WC ASPHALT WEARING COUF lOOmm BTB RECYCLED ASPHALT BAS

(2x50 mm)

UPPER-

SUBBASE

C4

{HARD CEMENTED TOP 50

SOFT UNCEMENTED lOOn

FIGURE 3.42

HORIZONTAL PUMPING OBSERVED IN THE TEST PIT, BOOm".

AWAY

FROM SECTION 2

E

~ 600

ll:

:>

4100

~

" u o

SURFACE

DISTRESS

~H SURFACE fMPERATURE

-

~ 800 -

Z o

800 -

~ 4100

~ 200

•••

80 100

200 300 3150

REPET IT IONS AT 70llN DUAL WHEEL LOAD 1.10· I

( 0 )

SECTION

J

500

u

...

1&1

0

0

III

!

i

!

Z

52 l-

4100

~

SO

300 ul/l

••.

0_

!

200

~!

100

Q c

II:

0

1,15

0,5

0

(b)

70llN

TRAFFICKING DUAL WHEEL LOADS

100llN

SURFIICE

DISTRESS rURFACE

WATER

SO

100 150

200

2150

300

3150

NUMBER OF REPETITIONS ,.

10

3)

SECTION 2

~ SUBSURFACE

WATER

4100

.1\

8

"8

AVE

A

4150

500

FIGURE 3.43

AVERAGE ROAD SURFACE DEFLECTION AND

RADIUS OF CURVATURE UNDER A 40kN DUAL

WHEEL LOAD ON SECTION I AND SECTION 2

The average RSD and RC under a 40 kN dualwheel load on

Section 2 are illustrated in Figure 3.43 (b)• The initial

RSD and RC were 0,44 nun and 230 m, respectively.

These values were both better than those measured on Section 1, although it is also relatively low compared with the previous structures at Mariannhill, Figtree and umgababa.

The

RSD increased to approximately 0,80 nun and the RC decreased to almost 80 m when fatigue cracks appeared on the surface of the section.

This "end state" compares favourable with those found on Section 1.

During the application of water, the effect was not rapid changes in these indicators, (RSD and RC) mainly because of the originally poor state of the structure.

Part B of the section (surface and subsurface water) experienced the highest RSD (approximately 1,25 nun) and the lowest RC (approximately 50 m) at the end of the test.

This section changed also from flexible (F) to very flexible (VF).

The slight decrease in RSD and increase in

RC at the end of the test on part A of the section is believed to be partly due to difficulty in measurements

(irregular surfaces) and interlocking* of the asphalt blocks

(150 nun) during trafficking.

It was noted during the trafficking that less rocking of the individual blocks occurred during repetitions 300 000 to 400 000.

The netto effect however on the total section (part A and part B) was that the rocking of the blocks increased to the end of testing.

The average MDD depth deflections under a 40 kN dualwheel load at various stages of trafficking on Section 1 are illustrated in Figure 3.44(a).

The figure indicates relatively high depth deflections, which was also reflected in the relatively high RSD values.

According to this result, approximately 50 per cent of the RSD resulted from the

Similar to those found with interlocking block paving (Clifford,

1985).

0,2 o EF LEe TlON ,•••• I

0,4

0

0

'0

100 -

ISO

200

E

E

II:

:>

Q

••

..•.

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II:

..

no

300

3'0

400

4'0

..

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J:

'00

,SO •

600

••

6'0

J:

Ia.

100

••

800

900

.--~_.,---

••

-N"

10

58800

-"-10517a

H" 315121

}

TRAffICKING

LOAD

70 11."

II

"d

",,

,

;

'"

,

,"

"',

'

",

"

'

,

,

M'

NRfPETITIO~~S

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AT uN

•.•• !o88()()

1,2,,0 •••

WEA't'NG

COUflS£

100 "" ••

!tEeye

LED

BASE

ISO m.,

UPPER

SU8BA 5 E

150 mm

LOWER

SUB B ASf

SELE(TE

SU8GRADE

0

0

0 eo

1411

0,2

..

III

J:

300

I

..

0

0

II:

•.

••

4'0 u

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J: la.

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REPETITIONS su"'ACf

DI5TflES'S ON

AT '0 1000

AVERAGE MULTI-DEPTH DEFLECTION UNDER A 40kN DUAL WHEEL LOAD

AT VARIOUS STAGES OF TRAFFICKING ON SECTIONS I AND 2

W

•...

0

0'

selected subgrade layer, downwards.

The other 50 per cent resulted from the surfacing, base and the two subbase layers.

Relatively high deflections (0,15 nun to 0,20 nun) were measured in the upper subbase (145 nun to 300 nun) and the selected subgrade (450 nun to 640 nun).

Only slight variations in deflection were measured with increase in traffic and it is believed that the initially poor state of the upper subbase contributes largely to these high deflections.

A relatively small relative deflection (0,10 nun) was measured within the lower subbase layer, which was relatively strongly cemented.

The individual contributions to the RSD from the different layers are given in Table 3.20.

The table indicates increases in contribution from the asphalt and selected subgrade layers, and decreases in the contributions of the two subbase layers.

Not much difference were noted between the state at 58 800 repetitions, and the end of the test.

The results also indicates that initially 35 per cent of the

RSD originated from within the upper subbase.

This value decreased to 16 per cent during the first 58 800 repetitions.

At the end of the test, only 14 per cent of the RSD originated from this layer owing to possible compaction under HVS traffic.

The contribution from the selected subgrade layer increased from 45 per cent to 52 per cent and these of the asphalt layers from 3 to 22 per cent.

This relatively high increase in relative deflection probably indicates a weakening of the recycled asphalt base layer during trafficking.

The average MOD for Section 2 under a dualwheel load of 40 kN at different depths below the road surface for various stages of trafficking is illustrated in Figures 3.44(b).

Beginning of test

At 58 800 reps.

(Fatigue distress)

End of test

Surface and base course

Upper subbase

Lower subbase

Selected subgrade

3%

35%

17%

45%

22%

16%

10%

52%

25%

14%

9%

52%

The figure also indicates that marked increases in the deflections of all the layers during the first 70 000 repetitions of the 70 kN wheelload occured.

The RSD measurements along the centre line of the section indicate that the

RSD increased from point 4 to point 12 on this section caused by the higher subgrade deflections experienced at measuring point 12.

There was however no clear indication that fatigue distress started earlier at the vicinity of point 12 than on the rest of the section.

This could be an indication that the fatigue crack development in this case is rather a function of the upper subbase support than the subgrade support.

The low RC also indicated that inadequate support resulted from the upper subbase.

In Figures 3.46(a) and (b) the initial and final MOD deflections on part A (MDD4) and part B (MDD12) are illustrated, respectively.

Both figures indicates further weakening of the upper subbase.

A relatively marked increase in the deflections of all the layers at MDDl2 (surface and subsurface water) was experienced.

It is believed that there are mainly two reasons for the higher deflections measured with MDDl2.

Firstly the initial deflections at MDD12 were

900

E

::L

800 z

...

700

U

III

J

...

600

III

0

500

400

1400

1300

1200 llOO

1100

-t:--{t

MOD4 MD08 MOOI~

.-e-- -

---e--.-e--

~I I I

012345618

I " I I I I 1.,

9 lOll 1213141516 MEASURING

TEST SECTION 2 POINTS

300

200

MOO MOD

MOO

100

0

2

1

1 1

3 ~

5 6

7

®

9 10 II

@

13 14

LONGITUDINAL MEASURING POINTS ALONG SECTION

FIGURE 3.45

ROAD SURFACE DEFLECTIONS UNDER A 40kN DUAL WHEEL LOAD ON THE CENTRE LINE

DURING VARIOUS STAGES OF TRAFFICKING ON SECTION 2

4~0 i

&00

!

III u

:

1140

Q

III

:z:

,.

..

It:

C 1000

Q

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••...

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~~:.

~'T~~~GLY

C[MUITED

INITIAL AND FINAL MULTI-DEPTH DEFLECTIONS UNDER A 40kN' DUAL WHEEL

LOAD AT M[){)4 (F?1RTA)AND MDDI2 (PARTB)

higher from the start of the test.

Secondly, it is believed that the subsurface water played a major role in decreasing strength of especially the granular upper subbase, therefore more support was needed from the lower layers, resulting in higher subgrade and hence total deflections.

As was done with the MOD deflection results at the previous three sites, mechanistic analyses were also done on the results at this HVS site.

In addition vertical stress analyses indicated that the highest stresses occurred within the recycled asphalt base layer and increase markedly with increase in wheel load (De Beer, 1985(b».

This observation is believed to be indicative of the severe stress conditions that the asphalt experienced during trafficking on this structure (weak upper subbase support).

Both fatigue distresses (bigger and smaller blocks - see Section 3.5.2.1) in the asphalt are believed to be directly a result of this condition.

It is however appreciated that the".linear

elastic model is based on a few important assumptions such as homogeneitYi isotropy etc., nevertheless this rather simple modelling can be used to indicate very important behavioural characteristics.

The different effective elastic moduli values for the three test sections are given in Tables 3.21, 3.22 and

3.23.

The moduli values were calculated at 40 kN wheelload at 520 kPa tyre pressure.

The modulus of the asphalt wearing course in this case was kept constant at 3 000 MPa during the analyses.

This was done because no relative deflections were measured in this layer.

The results in the tables indicates very low moduli values for both weakly cemented subbase layers.

On Section 1 (Table 3.21) the modulus values ranging from 32 MPa to 55 MPa.

On

Section 2 (Table 3.22) the values ranging from 46 MPa to

TABLE 3.21

Effective elastic moduli for the different pavement layers for various stages of traffickinq at 40 kN dual wheel load on Section 1

NwDber of repetitions

With a dual wheel

Load of 70 kN (N)

Wearing course*

(MPa)

Average values for points 4, 8 and 12:

10

58 800

261 009

333 847

3 000

3 000

3 000

3 000

Base

(MPa)

1 563

1 206

1 228

684

Upper subbase

(MPa)

38

35

32

55

Lower subbase

(MPa)

55

51

51

54

Selected layer

(MPa)

72

27

27

29

Subqrade

(MPa)

182

123

139

145

TABLE 3.22

Effective elastic moduli for the different pavement layers at various stages of trafficking at 40 kN dual wheel load on Section 2

Number of repeti tioDS

(N)

(70 and 100 kN)

Wearing

Course

(MPa)

Base

(MPa)

Upper

Subbase

(MPa)

Lower

Subbase

(MPa)

Selected

Layer

(MPa)

Subqrade

(MPa)

Average values for points 4, 8 and 12:

10

10 000

100 000

211 460

301 488

3 000

3 000

3 000

3 000

3 000

396 979 (EPWP)

3 000

446 249: for points 4

8 (EPWP)

12

3 000

3 000

3 000

1 490

1 236

1 019

847

787

572

876

18

679

89

75

58

48

46

120

44

59

24

115

95

92

68

73

66

123

71

43

55

36

26

19

17

15

39

25

12

133

106

74

60

52

44

51

85

44

TABLE Effective elastic moduli at 40 kN dual whee110adfor the different layers at various stages of trafficking on Section 3

Number of

Repetitions

(N)

Wearing Base

Course

Upper Lower Selected Subgrade

Subbase Subbase Layer

10 000

216 023

713 114

866 379

996 813

3 000

3 000

3 000

3 000

3 000

1 245 152

1 235 135

1 186 106

476 136

71 167

238

198

179

180

232

122

111

98

85

115

115 MFa and on Section 3 the values ranging from 106 MFa to 238 MFa.

Comparing these moduli values to those of the previous three sites it is very low.

The reason is believed to be the initial state of especially the upper subbase which was mainly granulated approximately equivalent to G4 (EG4) or lower quality material.

The granulated state of this layer was confirmed also after HVS testing when inspection test puts were made.

The moduli values of the upper subbase at Section 1 are the lowest of the three sections.

This is believed to be due to the relatively strongly cemented SO mm to 75 mm layer found in the structures of Sections 2 and 3.

In the case of the rather low moduli values of the lower subbase, which was also relatively strongly cemented, it is believed that the MOD levels during instrumentation were not exactly positioned at the layer interfaces because this is virtually impossible from a practical point of view.

Because of the relatively weakness of the upper subbase (or selected layer) higher relative deflections were measured in the lower subbase hence the lower

moduli values.

It is however believed that the moduli values of the upper subbase are indicative of its relatively weak and granulated state.

This weak upper subbase can therefore be viewed as a weak interlayer between the asphalt and the relatively strongly cemented lower subbase.

The effect of interlayers in this type of pavement structure is however analysed in

Chapter 5.

Using the modular ratios of the asphalt and upper subbase, which is an indication of balance* in the structure.

If the modulur ratios of the initial moduli values of the base and upper subbase are calculated and compared with the sum of the horizontal tensile strains (£t)

**

at the bottom of the treated (asphalt and cemented) layers, a distinct difference in the results at Van Reenen's Pass and the other HVS tests at Mariannhill, Figtree and

Umgababa is obtained.

See Figure 3.47.

The figure indicates that a very high modular ratio (E1/E2) of approximately 40 resulted in a relatively high Fdp.

Simply this is indicative that the potential for fatigue distress in this structure (mainly the asphalt layers) is very high and that relatively early fatigue cracking

is

expected at the Van Reenen's Pass structure.

In Chapter 5

it is shown that the thicker the interlayer and the weaker

(wet and granulated vs cemented) the higher the Fdp.

The higher this ratio the poorer the "balance", the higher the fatigue distress potential, Fdp.

The scalar quantity of the sum of the maximum horizontal tensile strain at the bottom of the treated layers in the structure is called the "Fatigue distress potential", Fdp, of the structure under consideration (See Chapter 5, Section 5.2.1.2).

E, ' INITIAL.

EFFECTIvE ELASTIC MODUL.US OF THE ASPHALT BASE L.AYER

E1' 'NIT IAL. EFFECTIVE ElASTIC MODULUS OF THE WEAKL.

y

CEMENTED UPPER

SUBBASE L.AYER

} 40 ~N ;

$20 ~Po

FclP

'!

"

,.,

WHE RE i" : ASPHALT BASE

2 UPPER SuBBASE

3 . LOWER SuBBASE

VAN REENEN" PASS.

N3/6

(EARL. Y FATIGUE DISTPESS OF AsPHALT

IN DRY STATE)

20

10

..J

..J

ii:

<:%

:E

% z z c z

...,

'-

0 500 1000

FATIGUE DISTRESS POTENTIAL, Fdp

(I-'-()

FIGURE 3.47

RELA TIONSHIP O~ THE INITIAL EFFECTIVE ELASTIC MOOULAR

RArlO (E// E2) BETWEEN THE ASPHALT BASE ANO WEAKLY

CEMENTEO UPPER SUBBASE LAYER ANO·THE FATIGUE

OISTRESS POTENrtAL(Fdp)AT THE FOUR OIFFERENT rl VS

SITES.

- 3.116

-

TABLE to have inadequate subbase support

*Locality

(Jail)

Lacroix

(Slow lanes) deflection

(1982) emm)

Northbound :

28,5 - 29,5

33,7 (fill)

35,2

36,2 - 36,4

37,5

43,5

46

Southbound

29,7

30,S

33 - 33,4

36

39,4

46,3

> 0,4

> 0,4

> 0,4

'V

0,2

'V

0,2

'V

0,4

> 0,4

> 0,4

> 0,4

> 0,4

> 0,4

> 0,4

> 0,4

DCP measurements

(lIIIII!blow)

2,5 - 5

2,5 - 5

2,5 - 5

1,0 - 2,5

2,5 5

2,5 - 5

2,5 - 5

2,5 - 5

Likely condition

Slope instability

Weaksubbase

Weaksubbase

Weakupper subbase

Weakupper subbase

Weakupper subbase

Slope instability

Weakupper subbase

Weakupper subbase

Weaklower subbase of renumberinq of this route, the kilometer distances chanqed (increased) by approximately 9,8 km.

In order to evaluate the validity of the HVS findings on this site certain predictions for fatigue distress of the asphalt layers were made and was re-evaluated during March

1985.

Because this was a rehabilitation project and not a pre-planned HVS experimental section, the HVS findings were used to predict the behaviour of the total road using also the DCP and Lacroix survey results.

Table 3.24

lists pavement areas likely to have inadequate subbase support according to the evaluation performed in

1983.

Lacroix deflections and DCP measurements were used

to identify these areas.

Deflection levels on these areas were mainly above the 0,4 mm given as a warning level in draft TRH12 (NITRR, 1983).

As slope instabilities could well account for some of these high deflections, DCP measurements were used to determine the condition of the weakly cemented subbase layers.

It was stressed that slope instability cracks should be sealed immediately after they had occur to prevent the ingress of water into the subbase layers and the failure planes of the unstable slopes.

After the HVS, deflection and DCP results had been analysed, certain conclusions and recommendations were made.

A summary of these findings is given below:

The need for major rehabilitation within five years of testing is not foreseen provided timely sealing, local repairs and subdrainage improvements were undertaken.

The expected behaviour and likely need for future sealing is given in Table 3.25.

TABLE 3.25 - Expected behaviour of rehabilitated pavement

(De Beer, 1985(b))

Percentage of road affected

Predicted appearance of fatigue cracks :

10 - 15

85 90

Sealing likely to be needed before

'V

'V

1988

1995

If not sealed in time critical deformation

(20 mm) is likely to occur owing to ingress of:

Surface water in

Surface and subsurface water in

'V

1990

'V

1997

'V

1988

'V

1995

It is recommended that the pavement be inspected periodically, especially after rainy seasons, to identify the areas with inadequate drainage.

The behaviour of the asphalt layers subbase layers in this pavement structure

is

a function of the support from mainly the especially the upper subbase.

Where the support

is

adequate, (cemented) the relatively rigid recycled asphalt layer is expected to be able to carry the predicted future traffic without the appearance of serious fatigue distress.

The investigation has shown that the asphalt produced with a 70/30 ratio of reclaimed bitumen to virgin material experienced early distress cracking when there was inadequate support.

Total structural and functional failure occurred when water entered this weak layer.

Further in this paragraph a summary is given of the method of predicting when surface cracking (fatigue distress) may be expected on this route.

Two predictions were made regarding the time (in years) expected to elapse before fatigue distress and pumping occured, viz: for the weaker 10 to 15 per cent and for the remaining 85 to 90 per cent of the route.

This elastic simulation previously mentioned was used to calculate the effective horizontal strain values (microstrain) at the bottom of the asphalt base layer.

The expected fatigue life to cracking (visible at surface) on the three sections is given in Table 3.26.

The initial effective elastic moduli of the asphalt base layer varied between 1 000 and 2 000 MFa.

The effective elastic moduli of the upper subbase varied between 30 and 40 MPa on Section 1 and between 45 and 150 MPa on Sections 2 and

3. (See previous Tables 3.21; 3.22 and 3.23)

TABLE 3.26 Expected fatigue life to cracking of asphalt base (De Beer, 1985(b»

Section

Dual wheel load

(kN)

(lJ£ )

Effective elastic moduli of the

Strain* asphalt base layer (MPa)

1 000 2 000

1

2 and 3

40

70

40

70

100

400

690

290

520

620

200 000

20 000

900 000

72 500

33 500

50 000

6 000**

160 000

16 000

8 250** where Pl

P2

=

TEST wheel load (kN)

=

STANDARD wheel load, 40 kN (E80) d

= damage exponent.

In order to convert the fatigue lives from Table 3.26 to standard E80s the following method was used:

-+-

-+-

(Pl)d

P2

=

Np2 ! NP1 d

=

..............

log (Np2!Npl)

........................

log (Pl!P2) from (3.4)

(3.6)

where Np2

=

Number of equivalent standard repetitions

{E80s}

Npl

=

Number of actual repetitions before cracking on the surface results from wheel load Pl.

The different damage exponents calculated from the data in

Table 3.26 are given in Table 3.27.

Effective elastic moduli of base layer

(MPa)

1

2 and 3

70

70

100

4,11

4,50

3,59

3,79

4,11

3,24

The results

in

Table 3.27

indicates that there is a relatively small variation

in

the exponent value, d.

Because the prediction would be for a number of full years, it was decided to use an average value, viz: 3,89, with a standard deviation of 0,45, for the damage exponent.

The number of equivalent standard axles required to produce surface fatigue distress, calculated with the average value, is given in Table 3.28.

1

2

3

TABLE 3.28 - Number of equivalent standard axles required to produce fatigue cracking visible on the surface of the asphalt

Section

Traffic loading (kN)

Actual repetitions

70

70

100

40

100

58 800

100 000

201 000

801 488

195 000

Standard repetitions

(E80s)

518 556

7,98 x 106

7,69 x 106

In order to calculate the number of years before cracking occurs (crack-free periods) the following equation was used (NITRR, 1985):

Where Ne

Ni fy

= cumulative equivalent traffic (E80s)

= initial traffic

= cumulative growth factor,

=

365 [(1

+ O,Oli) ((1 + O,Oli) y

-1

1]

(O,Oli)

With a compound growth rate of 10 per cent in E80s on this route the prediction (crack-free) period can be calculated.

The initial traffic was obtained from traffic weight classifier (TAWC) data obtained in 1982.

This data is given in Table 3.29.

TABLE 3.29 Traffic count and number of E80s calculated for this route

Direction of lane

Axle group weight

(tons)

Average count daily

(axles)

Daily count

Southbond 0 - 2

2 - 4

4 - 6

6 - 8

8 - 10

10 - 12

12 - 14

Northbound o -

2

2 - 4

4 - 6

6 - 8

8 - 10

10 - 12

12 - 14

3 708

445

519

237

95

13

0

TOTAL

3 886

475

716

480

106

9

0

TOTAL

1

10

83

141

150

45

0

430

1

10

115

286

168

31

0

611

(a) For 10 to 15 per cent of this route:

Section 1)

Ne

Ni i

=

518 556 E80s (from Table 3.28)

=

521 E80s per lane

=

10 per cent

-+-fy

=

365 (1,1) 1(1,1} y- 11/0,1

=

Ne/Ni From eq.(3.7)

Y

-+-(1,1) - 1

518 556

=

(521 x 4 015)

-+y

= log (1,25}/10g(1,1)

In this case the crack-free period from the time of

HVS testing is therefore two to three years.

(b) For the remaining 85 to 90 per cent of the route (HVS

Test Sections 2 and 3)

Ne

=

7,98 x 10

6

(from

Table 3.28) i

=

10 per cent

=

16,5 years

Ne

=

7,69 x 10

6

(from

Table 3.28)

For Sections 2 and 3 a conservative estimate (due to uncertainty about the long-term growth rate, i) is that the crack-free period will be longer than 10 years, probably between 10 and 15 years.

The expected behaviour of the rehabilitated pavement is summarized in previous Table 3.25.

This route was re-evaluated during March 1985.

According to Table 3.25 the prediction was made that fatigue cracking would appear during 1985 on some 10 to 15 per cent of the length of the route under consideration.

A visual inspection was undertaken to assess the validity of this prediction.

Before HVS testing a Lacroix deflection survey was done in the slow lanes of the route.

Deflection was measured every 5 m.

At the HVS site, however, a Benkelman beam deflection survey was done along the 100 m of test site, at 1 m intervals prior to HVS testing.

The various deflection results are illustrated in Figure 3.48(a), (b) and (c).

The position of the three test Sections are also indicated in the figure.

The average deflection of the site was 0,54 Mm.

(See Figure 3.48(c».

The crack survey in March 1985 revealed that the cracked areas coincides with the high deflection areas were cracked and signs of pumping were present.

The crack patterns were typical crocodile fatigue cracking, as experienced with the HVS in

1983.

The Lacroix deflection and

DCP results were used to identify possible "weak" areas in the subbases of this route.

As previously mentioned, Lacroix deflections were measured along the 24 km of route.

Deflections were measured every 5 m and the 95th percentile of each 100 m was calculated.

It is accepted that the Lacroix deflectograph tends to measure lower deflections than the Benkelman beam (Coetzee, 1985).

In this case the Lacroix deflection should be increased by approximately 20 per cent, plus a constant of 0,13 mm to be added to the

Lacroix deflections to match the deflection measured with

Benkelman beam.

The Lacroix deflections measured during

1982 are illustrated in Figures 3.48 (a) and (b).

The average deflection (Lacroix) of the HVS site is also

1.20

0,80

E

!

z

...

v

...

•••

Q

~

0.40

1.00

CllACKINI SLO~I INSTAIII.ITY

TIIAFFIC

DYNAIIIC

ASSOCIATED

(SlAI.ID)

CIIACKINI

CONE ~tNITIIOIIETEII

(IUIICH

,.111

AND ~UII~INe

IDC~1 1•••

(IIAIICH "'51

1.1••

II AUGUST 1.12) oNYENTIONAI.

I

ASPHALT lASE

I

I

I

I

I

I

I

,

'm

•• _

IIECYCLED

;l

~ HVS SITE

ASPHALT IA'I

1[i\W4&/~

0.20

o

27,0 29,0 31,0 33,0 35,0 43,0 45,0 47,0 49,0 51.0

37,0 39,0 41.0

DISTA NCE 1 kml

E

E

(0) z

2 lv

0,60

...

'"

Q

0,40

1,20

1,00 o,eo

I

I

I

I

I

I

DEFLECTION, DCP AND CRACKING SURVEYS ON THE SOUTHBOUND CARRIAGEWAY

I

.-

~

SLOPE INSTAIILITY CIIACKINli l'EALED) IIIAIICH •• IS)

T1UFFIC ASSOCIATED CIiACKIHQ AND ~UIIPINe (IIAIiCH •• 151

INORTHI

I

I

I

I

I

I

I

I

DYNAMIC CONE ~ENETIiOIlETEIi (OCPlC ••• I ., •• ) IAUliUST "121

I

I••

CONYENTION~~-C

I

ASPHALT lASE

IIECYCLEO AS~HAI.T

IAIE

-lit

It) U) W' 110

N

01

••>

••

+

I.

+~l~'

~A

r

~~

·1

I

I

I

I

I

1

I

I

HY' SITf DEFl.fCTloN

I

(AYEIIAliE)

I

~HY' SITI

I

0.20

37,0 39.0

41.0

DISTANCE Ikml

(b) DEFLECTION, DCP, AND CRACKING SURVEYS ON THE NORTHBOUND CARRIAGEWAY

_ TIIAFFIC A'SoCIATED CIIACKINe AND ~UII~IHli (IIAIICH

"'51

I

OYNANIC CONI ~INETIIOIIETERIDCPU ••••

I.I •• )

(AUliUST ••• 2)

__ ;~;i;*,-

(c)

DEFLECTION, DCP AND CRACKING SURVEYS AT THE HVS SITE

(NORTHBOUND CARRIAGEWAY)

FIGURE 3.48

RESULTS OF THE DEFLECTION, DCP AND CRACKING SURVEYS

DONE AT VAN REENEN'S PASS

illustrated in the figures, including the crack evaluation of March

1985.

As previously mentioned the evaluation includes the section of route

(27 km to

33,5 km) which consisted of conventional asphalt base material.

It is interesting to note that most of the locations where the Lacroix deflection exceeded a value of

0,35 mm

(0,55 mm Benkelman beam), were subjected to traffic associated cracking.

Most of those cracking were situated in the outer wheel path.

According to the prediction in

1983, 10 to

15 per cent of this route should be subjected to this type of cracking during the period

1985 to

1986.

See previous Table

3.25.

The re-evaluation indicated that almost 12 per cent of the length of this route already excracking and pumping).

A summary of the different percentages of cracking obtained are given in Table

3.30.

Section

(Base material)

Carriage way

(Direction of traffic)

Conventional asphalt

Conventional asphalt

Recycled asphalt

Recycled asphalt

TOTAL

North

South

North

South

Percentage cracking*

2,7

4,0

-4,2

1,1

12,0

It is also worth while to mentioned that the areas likely to have inadequate subbase support according to the prediction made in 1983 (see previous Table 3.24), coincided almost exactly with the cracked areas found in the current investigation.

This re-eva1uation also indicated that approximately

23 per cent of the length of this route is subjected to slope instability cracking.

Almost 100 per cent of these cracks were sealed.

It is, however, important to note that traffic associated cracking did appear at some locations of these cracks also. In general though, the two types of cracking occur independant, provided the slope instability cracks are adequately sealed in good time.

The results from Table 3.30 indicated that approximately

6,7 per cent of the cracking appeared on the area with the conventional asphalt base layer.

Furthermore it is interesting to note that the higher deflections measured in

1982 also occurred on these parts, of the route, especially on the southbound carriageway.

It is believed that the conventional asphalt contributes largely to this since it is more flexible than the recycled asphalt.

Ten different HVS tests were summarized in this chapter.

The tests were done in Natal, at four different sites with the same design.

The HVS tests indicate that the weakly cemented subbase layers experience fatigue cracking and fracturing.

However, if the subbases remain dry, neither excessive permanent deformation nor excessive asphalt fatigue distress is expected in the design period (12 ro 50 MEaos in 20 years) of these structures.

It is believed that this design will be able to carry more than 30 MEaOs in the dry state.

In the wet state (excess porewater pressure), the durability (erodibility) of the upper subbase governs the functional life of this design.

The tests indicate that durability requirements

should take precedence over the strength requirements of these fine-grained weakly cemented materials, (at least the upper subbase).

Furthermore it is important to police overloaded vehicles, since asphalt deformation may occur especially at relatively high ambient asphalt temperatures.

Any cracks should be sealed immediately and regular drainage maintenance should be done.

The weakly cemented weathered granite material, however,

is

adequate for the subbase layers

in these designs.

The Berea Red Sand, treated with three to four per cent lime, is not recommended for the upper subbase in this type of design (Umgababa tests).

It can be concluded that the function of the weakly cemented subbase layers in this type of pavement design are not only to protect the subgrade layers against excessive deformation but also and more important to protect the asphalt layer against fatigue distress.

It is therefore the author's belief that the upper subbase is one of the most important structural layers

in the pavement.

Investment should be in this layer because it is more economical from a rehabilitation point of view to repair only the base or surfacing layers.

With the aid of linear elastic modelling (mechanistic analyses in this case) very important behavioural characteristics could be quantified.

Of the most important parameters were surface and depth deflections, effective elastic moduli values, horizontal and vertical micro-strains and radius of curvature.

The introduction of artificial changes

in

the moisture state or elavated temperature under controlled conditions enabled the precise defining of the different failure mechanisms encountered at the different sites.

The importance of subsurface drainage to avoid EPWP conditions is again realized.

Quality control during the construction of weakly cemented layers is very important and proved to be the key factor for longevity of these layers.

Weak interlayers must be avoided

(Van Reenen, N3/6).

Stable fills must be provided before weakly cemented layers are build (Figtree, N2/24).

Most of the HVS tests indicates that weakly cemented subbase layers undergo cracking and dramatic changes in their effective elastic moduli values occur.

From this it appears that the post cracked phase are very important because most of these layers were cracked during most of the testing.

In the following chapter an attempt is made to illustrate an example how to quantify weakly cemented layers during the postcracked phase.

It is considered only as an example, but certain principles used are believed to be very important when doing similar analyses.

OTTE, E (1978).

A structural design procedure for cementtreated layers in pavements.

DSc thesis, University of

Pretoria, South Africa.

FREEME, C R and STRAUSS,

J

A (1979).

Towards the structural design of more economical pavements in South Africa.

Proc.

3rd Conf. on Asphalt Pavements for South Africa, Durban.

FREEME, C R and WALKER, R N (1984).

Economic design of bituminous pavements.

Proc.

4th Conf.

on Asphalt Pavements for South Africa, Cape Town.

FREEME, C R (1984).

Sy!posium on: Recent findings of Heavy

Vehicle Simulator testing.

ATC 1984, NITRR, South Africa.

DE BEER, M (1984(a» .

HVS testing at Mariannhill, N3/1:

Detail Report.

NITRR Technical Report, RP/11, CSIR, Pretoria.

DE BEER, M (1984(b)).

HVS testing of National Road N2/24 between Illovo and Umgababa in Natal (Figtree).

NITRR Technical Report, RP/10/84, CSIR, Pretoria.

DE BEER, M (1985(a».

HVS testing of National Road N2/24 at umgababa in Natal, NITRR Draft Report TP/40/85, Pretoria,

CSIR,1985.

(Unpublished).

DE BEER, M (1985(b))• HVS testing and evaluation of the recycled base section at Van Reenen's Pass, N3/6 Upgraded detail report (Unpublished).

NITRR Technical Report,

TP/135/85, CSIR, Pretoria.

Standards for road construction materials.

Pretoria, NITRR.

RUST, F C (1985).

Load associated crack movement measurements during HVS testing.

NITRR Technical Note TP/68/85, Pretoria,

CSIR, 1985.

VAN ZYL, N

J

W and MAREE,

J

H (1983).

The behaviour of a high-standard crushed-stone base pavement during a heavy vehicle simulator test.

NITRR Technical Report, RR358 , CSIR,

Pretoria.

VILJOEN, C ELand VAN ZYL, N

J

W (1983) The "Marvil" Permeability Apparatus for in situ testing of surfacing and base coarse layers. Technical Note, TP/181/83, NITRR, Pretoria.

KLEYN, E G and VAN HEERDEN, M

J J

(1983)• ings to optomise pavement rehabilitation.

toria.

Using DCP sound-

ATC, CSIR, Pre-

MAREE,

J

H (1982).

Aspekte van die ontwerp en gedrag van padplaveisels met korrelmateriaal kroon1ae, D.Sc.

thesis,

University of pretoria, Pretoria.

DE BEER, M (1985(c» • Erosion test : Intermediate revised method and first results. Technical Note, TP/6/85, Pretoria,

NITRR.

NETTERBERG, F (1984).

Rapid Field test for carbonation of lime or cement treated materials.

RS/2/84, NITRR, Pretoria.

FREEME, C R, FRANCIS, V C, VILJOEN, A W, and HORAK, E.

(1982).

The Impetus of Heavy Vehicle Simulator Testing in Natal.

Proceedings of the Annual Transporation Convention, August

~, Pretoria, South Africa.

CLIFFORD,

J

M.

Some aspects of the structural design of segmental block pavements in southern Africa, D.Ing. Thesis,

University of Pretoria, 1984.

Bituminous pavement rehabilitation design,

CSIR, Pretoria.

NATIONAL INSTITUTE FOR TRANSPORT AND ROAD RESEARCH.

(1985).

Structural design of interurban and rural road pavements

(TRH4).

Pretoria, CSIR.

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